table of contents

TABLE OF CONTENTS Serial No 1 1.1 1.2 1.3 1.4 1.5 1.6 2 2.1 2.2 2.3 2.4 2.4.1 2.4.2 2.4.3 2.5 2.5.1 2.5.2 2.5.3 2.5.4 2.5.5 2.5.6 3 3.1 3.2 3.3 3.3.1 3.3.2 3.4 3.5 4 4.1 4.2 4.2.1 4.2.2 4.3 4.4 4.5 4.6

Title CHAPTER 1 : INTRODUCTION GENERAL BEHAVIOR OF BRIDGES UNDER EARTHQAUKES PRINCIPLES OF SEISMIC DESIGN STUDY BACKGROUND OBJECTIVE AND SCOPE THESIS OUTLINE CHAPTER 2 : REVIEW OF LITERATURE GENERAL DAMAGE DUE TO SEISMIC ACTIVITY DEVELOPMENT OF SEISMIC CODES SEISMIC VULNERABILITY ASSESSMENT EXPERT OPINION BASED ANALYSIS EMPERICAL METHOD ANALYTICAL METHOD RETROFIT STRATEGIES FOR BRIDGES CONCRETE JACKETS STEEL JACKETS PRECAST CONCRETE SEGMENT JACKET FRP STRENGTHENING EXTERNAL POST-TENSIONING SMART MATERIALS CHAPTER 3 : STRENGTH AND DUCTILITY OF BRIDGE BENT GENERAL DESCRIPTION OF THE HIGHWAY BRIDGE AND THE BENT ASSESSMENT OF BRIDGE BENT PLASTIC HINGE LENGTH AND ULTIMATE DISPLACEMENT OF PIER LATER STRENGTH AND DUCTILITY OF BRIDGE BENT SAFETY EVALUATION OF THE BENT SUMMARY CHAPTER 4 : SEISMIC PERFORMANCE ANALYSIS OF THE BEND GENERAL MODELING OF THE BRIDGE BENT PHYSICAL MODEL OF THE BENT ANALYTICAL MODEL OF THE BENT EARTHQUAKE GROUND MOTION TIME HISTORY ANALYSIS OF BENT ASSESSMENT OF BENT DAMAGE STATE SUMMARY iv

Page No 01 01 01 04 07 07 07 09 09 09 17 19 20 20 22 22 23 23 26 27 29 29 31 31 31 34 34 44 59 63 64 64 64 64 65 67 69 89 90

5 5.1 5.2 5.3 5.4 5.5 5.6 6 6.1 6.2 6.3

CHAPTER 5 : FRAGILITY OF RETROFITTED CONCRETE BRIDGE BENT GENERAL GROUND MOTION FOR INCREMENTAL DYNAMIC ANALYSIS CHARACTERISTICS OF DAMAGE STATE INCREMENTAL DYNAMIC ANALYSIS FRAGILITY CURVE DEVELOPMENTS SUMMARY CHAPTER 6 : CONCLUSION GENERAL CONCLUSION LIMITATION OF THE STUDY

v

92 92 92 96 97 97 104 105 105 105 106

Serial No

17

FIGURE Page No Figure1: A few damage scenario of a few highway bridges in the recent earthquakes 03 (Priestley et al. 1996) Figure 2.1: Nishinomiya-ko Bridge approach span collapse in the 1995 Hyogo-Ken 10 Nanbu earthquake. Figure 2.2: Flexural failure at the base of bridge pier during 1995 kobe earthquake 10 (Hanshin expressway). Figure 2.3: Weld failure of column longitudinal reinforcement, 1995 Kobe Earthquake. 11 Figure 2.4: Failed support columns and collapsed of the Cypress Street Viaduct (Loma 12 Prieta earthquake). Figure 2.5: Bridge collapse during the Northridge Earthquake in California. 13 Figure 2.6: Bond failure of lap splice at column base, 1989 Loma Prieta Earthquake. 13 Figure 2.7: Shear failure outside plastic hinging region, San Fernando Earthquake. 14 Figure 2.8: Brittle shear failure of column of the I-5/I605 seperator, 1987 15 Whittiereartquake Figure 2.9: Shear failure within plastic hinge region, 1971 San Fernando earthquake 15 Figure 2.10: Example of column shear failures, 1994 Northridge earthquake. (a) 16 Freeway atFairfax/Washinton undercrossing; (b) I-118 Mission/Gothic undercrossing Figure 2.11: Flexure-Shear failure at pier mid-height of Route 43/2 overpass, due to 17 premature termination of longitudinal reinforcement; 1995 Kobe earthquake. Figure 2.12: Concrete Jacket 23 Figure 2.13: Typical steel jacket retrofit details: (a) full height (b) typical section. 24 Figure 2.14: Effect of confinement on concrete, adapted from Priestley et al. (1996). 25 Figure 2.15: Steel Jacket Retrofit at Expressway in 1989, which was effective during the 26 1995 Kobe earthquake Figure 2.16: Precast concrete segment jacket 27

18 19 20 21 22 23 24 25 26 27 28 29 30 31 32 33 34 35

Figure 2.17: Before retrofit Figure 2.18: After retrofitted Figure 2.19: FRP strengthening of concrete pier Figure 3.1: 3D View of Bahadarhat Flyover Figure 3.2: Geometrical Model of Flyover Figure 3.3: Numerical evaluation of flexural and shear component of displacement Figure 3.4: Moment-Curvature Relationship of Pier 1 Figure 3.5: Moment-Curvature Relationship of Pier 2 Figure 3.6: Moment-Curvature Relationship of Pier 3 Figure 3.7: Moment-Curvature Relationship of Pier 4 Figure 3.8 : Moment-Curvature Relationship of Pier 5-21 Figure 3.9: Moment-Curvature Relationship of Pier 22 Figure 3.10: Moment-Curvature Relationship of Pier 23 Figure 3.11: Moment-Curvature Relationship of Pier 24 Figure 3.12: Pushover analysis of the bent Figure 3.13: Force-displacement Relationship of Pier 01 Figure 3.14: Force-displacement Relationship of Pier 02 Figure 3.15: Force-displacement Relationship of Pier 03

1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16

vi

27 27 29 32 33 35 36 37 38 39 40 41 42 43 45 46 47 48

36 37 38 39 40 41 42 43 44 45 46 47 48 49 50 51 52 53 54 55 56 57 58 59 60 61 62 63 64 65 66 67 68 69 70 71 72 73 74 75 76 77

Figure 3.16: Force-displacement Relationship of Pier 04 Figure 3.17: Force-displacement Relationship of Pier 05-21 Figure 3.18: Force-displacement Relationship of Pier 22 Figure 3.19: Force-displacement Relationship of Pier 23 Figure 3.20: Force-displacement Relationship of Pier 24 Figure 3.21: Ductility Evaluation of Bent Figure 3.22: Spectral Acceleration Figure 3.23: Spectral Acceleration (percentile and BNBC response spectrum) Figure 4.1: Geometry of Span of the Flyover Figure 4.2: Analytical Model of Bent Figure 4.3: Menegotto-Pinto steel model Figure 4.4: Fiber Section of Reinforced Pier Figure 4.5: Beam element for pier modeling Figure 4.6: Fiber Section of Reinforced Pier cap Figure 4.7: Time history graph of EQ-1 Figure 4.8: Pier displacement due to EQ-1 Figure 4.9: Time history graph of EQ-2 Figure 4.10: Pier displacement due to EQ-2 Figure 4.11: Time history graph of EQ-3 Figure 4.12: Pier displacement due to EQ-3 Figure 4.13: Time history graph of EQ-4 Figure 4.14: Pier displacement due to EQ-4 Figure 4.15: Time history of EQ-5 Figure 4.16: Pier displacement due to EQ-5 Figure 4.17: Time history of EQ-6 Figure 4.18: Pier displacement due to EQ-6 Figure 4.19: Time history of EQ-7 Figure 4.20: Pier displacement due to EQ-7 Figure 4.21: Time history of EQ-8 Figure 4.22: Pier displacement due to EQ-8 Figure 4.23: Time history of EQ-9 Figure 4.24: Pier displacement due to EQ-9 Figure 4.25: Time history of EQ-10 Figure 4.26: Pier displacement due to EQ-10 Figure 4.27: Time history of EQ-11 Figure 4.28: Pier displacement due to EQ-11 Figure 4.29: Time history of EQ-12 Figure 4.30: Pier displacement due to EQ-12 Figure 4.31: Time history of EQ-13 Figure 4.32: Pier displacement due to EQ-13 Figure 4.33: Time history of EQ-14 Figure 4.34: Pier displacement due to EQ-14 vii

49 50 51 52 53 55 60 61 65 65 66 66 67 67 69 70 70 71 71 72 72 73 73 74 74 75 75 76 76 77 77 78 78 79 79 80 80 81 81 82 82 83

78 79 80 81 82 83 84 85 86 87 88 89 90 91 92 93 94 95 96 97 98 99

Figure 4.35: Time history of EQ-15 Figure 4.36: Pier response due to EQ-15 Figure 4.37: Time history of EQ-16 Figure 4.38: Pier response due to EQ-16 Figure 4.39: Time history of EQ-17 Figure 4.40: Pier response due to EQ-17 Figure 4.41: Time history of EQ-18 Figure 4.42: Pier response due to EQ-18 Figure 4.43: Time history of EQ-19 Figure 4.44: Pier response due to EQ-19 Figure 4.45: Time history of EQ-20 Figure 4.46: Pier response due to EQ-20 Figure 5.1 Spectral acceleration of Earthquake ground motion records. Figure 5.2: Percentiles of spectral acceleration of earthquake ground motion records 5.2 Details of Retrofitted Technique Figure 5.3: Fragility curve development Figure 5.4: PSDM of As-build concrete pier Figure 5.5: PSDM of FRP Retrofitted concrete pier Figure 5.6: PSDM of Concrete Jacketed Retrofitted pier Figure 5.7: Fragility of as of concrete pier : Slight Damage Figure 5.8: Fragility of concrete pier : Moderate Damage Figure 5.9: Fragility of concrete pier : Extensive Damage Figure 5.10: Fragility of concrete pier : Collapse

83 84 84 85 85 86 86 87 87 88 88 89 94 95 99 100 100 101 102 102 103 103

Serial No 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24 25

viii

TABLE NAME Page No Table 1.1: Design Earthquake Ground Motions and Seismic performance of Bridges 05 (JRA 2002) Table 1.2: Relation between Complexity of Seismic Behaviour and Design Method 06 Table: 2.1: Damage Frequency Matrixes for Multi-Span Bridges 21 (Basoz and Kiremidjian, 1997) Table 2.2: Typical Mechanical Properties of FRP Laminates 28 Table 3.1: Geometries properties of the pier 33 Table 3.2: Material properties of the pier 34 Table 3.3: Plastic hinge length and ultimate displace of bent 44 Table 3.4: Lateral Strength of Bent 54 Table 3.5: Value of C_e in relation to effective height d of a pier section (JRA, 2002) 57 Table 3.6: Value of C_pt in relation to effective height d of a pier section (JRA, 2002) 57 Table 3.7: Failure mode: Flexural failure 57 Table 3.8: Failure mode: Shear failure of bridge pier 58 Table 3.9: Ductility of the Bent 58 Table 3.10: Earthquake ground motion 59 Table 3.11: Safety Evaluation of piers for, Z=0.15 62 Table 3.12: Safety Evaluation of piers for, Z=0.28 62 Table 3.13: Safety Evaluation of piers for, Z=0.38 63 Table 4.1: Characteristics of far field ground motion histories. 68 TABLE 4.2: Damage/limit state of bridge components (Hwang et al, 2001) 89 TABLE 4.3: Damage state of bridge pier 90 Table 5.1: Characteristics of the earthquake ground motion histories 93 TABLE 5.2: Damage/limit state of bridge components 96 (Hwang et al, 2001, Kartik et al 2012, Billah et al, 2012) Table 5.3: PSDM parameter for two type of bridge pier 100 Table 5.4: PSDM parameter for two type of Retrofitted bridge pier 101 Table 5.5: PSDM parameter for Concrete Jacketed Retrofitted pier 101

ix

ABSTRACT Elevated highways and bridges are very important elements of the infrastructure in modern societies. The performance of highway bridge systems observed in past earthquakes—including

ACKNOWLEDGEMENT

the 1971 San Fernando earthquake, the 1994 Northridge earthquake, the 1995 Great Hanshin earthquake in Japan, the 1999 Chi-Chi earthquake in Taiwan, the 2010 Chile earthquake, and the 2010 Haiti earthquake—have demonstrated that bridges are highly susceptible to damages

All praise goes to Almighty Allah who has given me the opportunity to do this thesis paper. It is a great pleasure for me to express my sincere appreciation to the Faculty of civil engineering

during earthquake causing a great economic loss. However, due to their importance, loss of

department, Chittagong University of Engineering and Technology for giving me the opportunity

functionality after a seismic event is not an acceptable performance criterion for the vast

to do this report. I want to mention the contribution of all those who have inspired, influenced and

majority of those structures.

guided me to complete this report successfully. First of all, I would like to express my heartiest gratitude to my venerable supervisor Prof. Dr. Md. Abdur Rahman Bhuiyan, Faculty of Civil

To this end, the current research focused on conducting reliability-based seismic performance

Engineering Department, Chittagong University of Engineering And Technology for his kind and

analysis of elevated highway and bridge bents which requires strong consideration of

sincere guidance, constructive criticism and personal supervision all through my work. I

performance evaluation and verification depending on different levels of design earthquake

wish to acknowledge my gratitude to all my respected teachers of the faculty, for their

ground motion according to the importance of them. In doing so, the damage scenarios of

suggestions and kind cooperation. Finally, I would like to extend my cordial compliments to

elevated highway/ bridges as observed in the previous earthquakes were critically reviewed and

my parents, friends and well-wishers for their inspiration and support throughout this report.

it was found that a large number of reinforced concrete elevated highway and bridges experienced severe damage/ collapse due to inadequate flexural and shear strength, and inadequate ductility of piers particularly in the Loma Prieta earthquake in 1989 and the Kobe earthquake in 1995.

As a case study, a typical elevated highway being constructed in

Bangladesh (i.e. multi-span simply supported elevated highway) was considered for seismic performance evaluation and verification. In this regard, the lateral strength, ductility and mode of failure of piers of the elevated highway were computed using the method of static pushover analysis. The sectional analysis method was employed to determine the flexural capacity at each section of pier. The seismic performance of the elevated highway pier was evaluated for a number of earthquake ground motion records using nonlinear dynamic analysis method. The peak ground accelerations (PGA) of the earthquake ground motion records considered in the analysis varied from 0.10g to 0.70g. The seismic responses of piers of the elevated highway were compared with damage limits available in literature and it has been found that the piers experience different levels of damage due to the design earthquake ground motion record being practiced in Bangladesh. Considering the damage levels of the piers of the elevated highway under the design earthquake a retrofitting strategy was arbitrarily selected with a view to improve the seismic performance of the piers. Finally, seismic fragility assessment of an as-built and a retrofitted elevated highway piers was carried out and it was found that seismic performance of the retrofitted pier can be significantly improved over the as-built elevated highway pier, under the design earthquake ground motion. iii

ii

CHAPTER 1 INTRODUCTION

because of high degree of redundancy generally inherent in building structural systems. Typically, bridges have little or no redundancy in the structural systems, and failure of one structural element or connection between elements is thus more likely to result in collapse. A

1.1 GENERAL

large number of highway bridges have experienced severe damage/collapse due to inadequate

Bridges are ubiquitous in today’s built environment, carrying highways through cities and

flexural strength and ductility of the bridge piers in the recent earthquakes. For instance,

countries and serving as the transportation lifeline of modern society. A Bridge is a structure

Figure 1(a) shows damage to the base of a column attributable to lap slice bond failure, which

providing passage over obstacle. The required passage may be for a road, a railway,

occurred in the Loma Prieta earthquake in 1989. Inadequate flexural strength may result from

pedestrians, a canal or a pipeline. It generates multifarious benefits for the people and

butt welding of longitudinal reinforcement close to the maximum moment locations. Figure

especially, promotes inter-regional trades and reduces traffic congestion and emergency

1(b) shows an example of flexural failure of a pier of the Hanshin Expressway in the Kobe

transit. Apart from quick movement of goods and passenger traffic by road and rail, it also

earthquake in 1995. This failure was initiated by failure of a large number of butt welds close

facilitated transmission of electricity and natural gas, and integration of telecommunication

to the pier base. Tension shift effects result in peak reinforcement strains being almost

links. Any damage in the bridge tends to halt the transportation system and it harms economy

constant for a height above the pier base equal to half of the pier diameter. The failure

devastatingly. The vitality of transportation networks must be ensured and the safety of the

depicted in Figure 1(b) was one of large number of piers in the Hanshin Expressway, where

users of transportation infrastructure must be guaranteed. Thus the importance of bridge

weld failure contributed to pier failure. Figure 1(c) and 1(d) show failure of flexural plastic

structures cannot be underestimated and bridges must be designed to adequately withstand the

hinges in the earthquakes. The low levels of transverse reinforcements were noted in the

force of catastrophic events.

piers. In Japan, a number of bridge piers developed flexure-shear failures at pier mid-height level during the 1982 Urakawa oki earthquake and the 1995 Kobe earthquake, as a

1.2 BEHAVIOR OF BRIDGES UNDER EARTHQAUKES Bridges are essential components of an overall transportation system as they play important roles in evacuation and emergency routes for rescues, first-aid, firefighting, medical services and transporting disaster commodities. The performance of highway bridge systems observed in past earthquakes—including the 1971 San Fernando earthquake, the 1994 Northridge earthquake, the 1995 Great Hanshin earthquake in Japan, the 1999 Chi-Chi earthquake in Taiwan, the 2010 Chile earthquake, and the 2010 Haiti earthquake—have demonstrated that bridges are highly susceptible to damages during earthquakes (Basöz et al. 1999, Yamazaki et

consequence of premature termination of longitudinal reinforcements. Figure 1(e) shows the flexure-shear failure of bridge piers initiated at bar cutoff locations around the pier mid-height level in the 1995 Kobe earthquake. Failure of eighteen piers of the collapse section of the Hanshin Expressway in the 1995 Kobe earthquake was also initiated due to premature termination of 33% longitudinal reinforcements at 20% pier height (Figure 1(f)) (Preistley et al. 2996). Shear failures also occurred extensively in the 1971 San Fernando earthquake, 1989 Loma Prieta earthquake, 1994 Northridge earthquake and 1995 Kobe earthquake (Priestley et al. 1996).

al. 2000).

Extensive damage of highway bridges triggered as a consequence of recent earthquakes has

Bridges give the impression of being rather simple structural systems. Indeed, they have

led to significant advances in bridge seismic design. Near-field ground motions developed in

always occupied a special place in the in the affections of structural designers because their

the Northridge and Kobe earthquakes are included in the 1996 Japanese and 1999 Caltrans

structural form tends to be a simple expression of their functional requirement. Bridges,

design codes (JRA 1996, Caltrans 1999). The conventional seismic coefficient method has

possibly because of their structural simplicity, have not performed well as might be expected

been replaced by ductility design method, and linear/nonlinear dynamic response analysis is

under seismic attack. In recent earthquakes in California in 1989, Japan in 1995, etc. modern

now used on routine basis in design of bridges with a complex structural response. This has

bridges designed specifically for seismic resistance have collapsed or have been severely

led to the development of performance based seismic design of highway bridge pier, which is

damaged when subjected to ground shaking of an intensity that has frequently been less that

being incorporated in various seismic codes (JRA 1996, 2002, Caltrans 1999, etc.)

corresponding to current code intensities. Earthquakes have a habit of identifying structural weakness and concentrating damage at these locations. With building structures, the consequences may not necessarily be disastrous, 1

2

(b) Flexural failure of a bridge pier in the 1995 Kobe earthquake, (c) Confinement failure at column top in the 1971 San Fernando earthquake, (d) Failure of flexural plastic hinges in bridge piers captured by connecting wall, Bull Creek Canyon Channel bridge, 1994 Northridge earthquake, (e) Flexural-shear failure at pier midheight of Route 43/2 overpass, due to premature termination of longitudinal reinforcement in the 1995 Kobe earthquake, and (f) Flexural failure above pier base of piers of the Hanshin Expressway in the 1995 Kobe earthquake

1.3 PRINCIPLES OF SEISMIC DESIGN (a)

(b)

Extensive damage of bridge structures in the recent earthquakes has exposed the deficiencies of the pre-1971 design rules specified by JRA (Japan Road Association) and AASHTO (American Association of State Highways Officials) in designing the bridge structures against earthquakes. Most of the existing bridges in the USA and Japan were constructed prior to 1970 (Kawashima, 2000). During this time period, insufficient consideration to the seismic forces was given, and bridge structures designed and constructed during this era have proved to be fully inadequate in the wake of earthquakes. Maximum 20% of the gravity force was considered for evaluating the seismic forces of the bridge structures; no other provision was specified in the AASHTO and JRA Specifications for highway bridges (Chen and Duan, 2000; Kawashima, 2000). A series of revisions to the specifications, especially by AASHTO and JRA, were accompanied by launching new guidelines/rules for specifying the seismic forces using the static lateral force method following the 1964 Niigata earthquake in Japan and the 1971 San Fernando earthquake in USA. However, some deficiencies in the revised

(c)

(d)

specifications were identified following the 1982 Urakawa-oki earthquake and the 1995 Kobe earthquake in Japan. In 1990 and 1995, the design specifications were revised again by incorporating the ductility design method in order to prevent the shear failure of the bridge piers. This method requires enough strength and ductility to ensure the structures to be capable of riding through an earthquake without collapse.

It is well known that the structural ductility is crucial in

dissipating seismic energy within the structures. An inelastic seismic design philosophy has been adopted in seismic design of bridge structures that incorporates formation of plastic hinges in different structural elements. Potentially devastating seismic forces are reduced by applying the inelastic design philosophy; however, the main flaw of this method is the necessity for the structure to experience some extent of damage. The origin of damaging (e)

(f)

effects of earthquakes on bridge structures is the unfortunate correlation between the

Fig.1. A few damage scenario of a few highway bridges in the recent earthquakes (Priestley

fundamental periods of vibration of major structures and the frequency content of the seismic

et al. 1996) (a) Bond failure of lap slices of bridge pier in the Loma Prieta earthquake in 1989, 3

input (Mullins, 1984; Priestley et al., 1996). 4

Extensive damage of bridges in the previous earthquakes as discussed earlier together with

earthquake in 1995 in Japan, it was a question of whether bridges should have higher seismic

research triggered as a consequence of recent earthquakes have led to significant advances in

performance than buildings. Aftermath of the Kobe earthquake it is now obvious that bridges

bridge seismic design. The conventional seismic coefficient method is being replaced by

should have higher seismic performance than standard buildings, because damaged buildings

ductility design method, and linear/nonlinear dynamic response analysis is being used on

cannot be restored if transportation is suspended due to collapse of bridges which have the

routine basis in design of bridges with a complex structural response. New treatment for

same seismic performance as buildings.

liquefaction and liquefaction-induced lateral ground movement was included in Japanese

The seismic performance of bridges shall be verified by a proper method in accordance with

Specifications, and verification of the ductility evaluation of reinforced concrete/steel

factors such as design earthquake ground motions, structure type and limit states. Seismic

single/frame columns is being conducted by various research organizations. Because of the

performance of bridges can be verified by static analysis and dynamic analysis according to

unsatisfactory performance of bridges in the 1995 Kobe earthquake, the Japanese Design

their dynamic characteristics. For example, for the bridges without complicated seismic

Specifications of Highway Bridges was revised in 1996. The code was further revised in 2002

behavior, the seismic performance can be verified using static methods; however, for the

based on the same main contents but including seismic performance criteria.

bridges with complicated seismic behavior, the dynamic methods of analysis can be employed for seismic performance verification. Table 1.2 summarizes the bridge types and

The performance based seismic design of bridges requires a strong consideration of

design methods applicable to seismic performance verification.

performance evaluation and verification depending on different levels of design earthquake ground motion according to the importance of the bridge. Moreover, the topographical, geological and soil condition, and site conditions, etc. should be taken into account in the seismic design of bridge (JRA 2002). Considering all these aspects, the entire bridge system capable of fully resisting earthquake forces shall be designed, by increasing ductility and strength of structural members. Necessary seismic performance shall be secured in the design of individual structural member of the bridge and the entire bridge system. Table 1.1 shows the different seismic performance in view of safety, serviceability and reparability for seismic design of bridge structures (JRA, 2002).

In verifying seismic performance, the limit state of each structural member should be appropriately determined in accordance with limit states of the bridge. For example, for seismic performance level 1, limit states of bridges are established so that the mechanical properties of the bridge are maintained within elastic ranges, for seismic performance level 2, these limits are established in such way that the structural member, which is allowed to experience plastic behavior, is allowed to deform plastically within a range of easy functional recovery, etc. This requires that the strength and ductility describing the capacity states of bridges should be properly evaluated. Table 1.1 shows the anticipated function of a bridge after a design earthquake. Important bridges undergo “limited damage,” in which “damage” does not exceed the stage in which the restoring force of main structural components initiates deterioration. Prior to the Kobe 5

6

1.4 STUDY BACKGROUND

In Chapter 4, Seismic vulnerability of the bridge bents under twenty far field earthquake

In the preceding sections, the seismic behavior, and state-of-the-art principles of seismic

ground motions was evaluated.

design of bridge structures have been discussed in a summarized way. From the discussion it In Chapter 5, Fragility functions are derived based on simulation results from nonlinear time

comes into view that:

history analysis, and then they are combined to evaluate the overall fragility of the bridge

1. In the previous moderate to strong earthquakes, a number of reinforced concrete piers suffered damage in different modes of failure, such as flexure, flexure-shear and shear modes of failure, as a consequence of inadequate

bents. Probabilistic Seismic Demand Model (PSDM) is employed to deprive the analytical curves. In developing the fragility curves displacement ductility of pier is considered as the Engineering Demand Parameter (EDP), and the Pick Ground Acceleration (PGA) is utilized as Intensity Measure (IM) for each ground motion record.

flexural strength, premature termination of longitudinal reinforcement and loss of confinement effect or combination of these effects. 2.

Finally, Chapter 6, presented the summary and conclusions attained from this research.

A good perceptive of seismic behavior of bridges is necessary towards a rational design of bridges for design earthquake ground motion.

3.

Seismic design of bridges requires a proper selection of earthquake ground motion and expected seismic performance during and after earthquakes.

1.5 OBJECTIVE AND SCOPE Based on the background summarized in Section 1.3, the present research was carried out to meet the following objectives:

1. To study the damage scenario of bridges as observed in the previous earthquakes 2. To estimate strength and ductility capacity of a typical bridge pier 3. To conduct seismic performance evaluation of a typical bridge pier subjected a series of earthquake ground motions

1.6 THESIS OUTLINE This thesis is arranged in six chapters. In the present chapter short preface and objective are presented. The content of the thesis is organized into the following chapters: In Chapter 2, a comprehensive literature review of bridge response during earthquake, retrofit and strengthening techniques of bridge are presented. In this chapter application of various techniques developed for bridge strengthening and their comparative performance also presented. In Chapter 3, Seismic performance of the bridge bent is evaluated. To evaluate capacity of the bent static pushover analysis and moment curvature analysis were conducted.

7

8

CHAPTER 2 REVIEW OF LITERATURE 2.1 GENERAL Highway bridges constitute a large portion of the national wealth and build up the foundation for infrastructure development. They perform as the arteries to establish link between cities and across country to provide a smooth and fast communication system, Bridge are very vulnerable to damage under major earthquakes (Basoz and Kiremidjian, 1999) such as the 1971 San Fernando Earthquake, the 1994 Northbridge earthquake, the 1995 Great Hanshin earthquake in japan and the Chi-Chi earthquake in Taiwan, the 2010 Chile earthquake and 2010 Haiti Earthquake, which extensively cause direct and indirect economic impact. In recent earthquakes in California, USA, Cobe, Japan and central and South-America, modern bridges specifically for seismic resistance have collapsed or have been severely damaged when subjected to ground shaking of an intensity that has frequently been less than that corresponding to current code intensities.

Figure 2.1: Nishinomiya-ko Bridge approach span collapse in the 1995 Hyogo-Ken Nanbu earthquake.

Many Existing bridges in Bangladesh may be inadequate in respect to the seismic

A 630m segment of the elevated Hanshin Expressway was collapsed in the Kobe earthquake.

performance required by the current codes and guidelines. Many of them were designed

The bridge consisted of 18 single circular columns monolithically connected to the deck and

without any earthquake resisting criterion, because they were built prior to earthquake

founded on groups of 17 piles. There were 18 spans in total, all of which failed. Both poor

resistance design codes ; Others were designed to resist horizontal actions but without the

structural construction and soil condition of the site are responsible for the flexural failure of

principals of the capacity design or are built at a site in an area where the seismic hazard has

the expressway (George et al, 2000).

been re-evaluated and increased. In order to upgrade the seismic performace of existing vulnerable RC structures various rehabitation techniques are available. The major techniques for structural rehabitation of RC bridge include encasing bents using steel, Fibre reinforced polymer (FRP) or reinforced concrete (RC) Jackets or by adding new structural elements.

2.2 DAMAGE DUE TO SEISMIC ACTIVITY Response of a bridge structure during an earthquake is likely to be influenced by proximity of the bridge to the fault and site conditions. Both of these factors affect the intensity of ground shaking and ground deformations, as well as the variability of those effects along the length of the bridge. Many of the sites were subject to liquefaction and lateral spreading, resulting in permanent substructure deformations and loss of superstructure support such as Nishinomiyako Bridge shown in figure 2.1.

Figure 2.2: Flexural failure at the base of bridge pier during 1995 kobe earthquake (Hanshin expressway).

9

10

Inadequate flexural strength may also result from butt welding of longitudinal reinforcement close to maximum moment locations. Figure 2.3 shows an example of flexural failure of a column of the Hanshin expressway in the 1995 Kobe earthquake, caused by failure of a large number of butt welds at the same location, close to the column base.

Figure 2.4: Failed support columns and collapsed of the Cypress Street Viaduct (Loma Prieta earthquake). Severe damage was caused during the Northridge earthquake during 1994 in California. Northridge earthquake occurred on a hidden fault northwest of downtown Los Angeles very near to the location of the Fernando earthquake.The significant movements resulted in a shear during the earthquake resulted in a shear failure of the reinforced concrete wing wall and shear keys at the abutment with subsequent loss of support of the superstructure and this followed by failure of the superstructure at the diaphragm over the continuous support, leaving the flared column standing (Denis et al, 1994) as shown in the figure 2.4.The majority of the bridge damages were due to shear failure of column or shear and flexural combined Figure 2.3: Weld failure of column longitudinal reinforcement, 1995 Kobe Earthquake.

failure mode (Denis et al, 1994).

During 1989 Loma Prieta earthquake,the Cypress Street Viaduct, a double-deck freeway section made of nonductilereinforced concrete that was suffered severe damage. Roughly half of the land the Cypress Viaduct was built on was filled marshland, and half was somewhat more stable alluvium (EERI, 1989). When the earthquake hit, the shaking was amplified on the former marshland, and soil liquefaction occurred. In an instant, 41 people were crushed to death in their cars. Cars on the upper deck were tossed around violently, some of them flipped sideways and some of them were left dangling at the edge of the highway. Figure 2.4 showesfailure of support columns and collapsed of the Cypress Street Viaduct.

11

12

Shear Failure occurred extensively in the 1971 San Fernando (Caltrans, Bridge design specification, 1993) 1994 Northbridge (EERI, 1995), and 1995 Kobe (Priestley et al, 1995) earthquakes. Figure 2.7 and Figure 2.8, from the San Fernando earthquake, is typical of brittle shear failure where flexural strength exceeds shear strength. There is no indication that plastic hinging developed at the member ends. In contrast, the column in Figure 2.7 and Figure 2.8, also from the San Fernando earthquake, has no apparent damage in the mid-region, but shear failure has clearly formed at the top of the column.

Figure 2.5: Bridge collapse during the Northridge Earthquake in California. Column longitudinal reinforcement was often lap spliced immediately above the foundation, with a splice length inadequate to develop the strength of the bars. Lap-splice lengths as short as 20 bar diameters were commonly provided that this is insufficient to enable the flexural strength if the column to develop (Chai et al, 1991). Figure 2.6 shows damage to the base of a column, attributable to lap-splice bond failure, which occurred in the 1989 Loma Prieta earthquake.

Figure 2.7: Shear failure outside plastic hinging region, San Fernando Earthquake.

Figure 2.6: Bond failure of lap splice at column base, 1989 Loma Prieta Earthquake.

13

14

Figure 2.8: Brittle shear failure of column of the I-5/I605 seperator, 1987 Whittier earthquake. The cause of failure of six of the seven bridge structures that failed in 1994 Northridge earthquake has been attributed to column shear failures (Priestley et al, 1994). Because of

Figure 2.10: Example of column shear failures, 1994 Northridge earthquake. (a) Freeway at Fairfax/Washinton undercrossing; (b) I-118 Mission/Gothic undercrossing

failure of transverse reinforcement, column shear failure often results in a loss of structural

In Japan, a number of bridge columns developed flexural-shear failures at column

integrity of the column, with subsequent failure under gravity loads (Figure 2.9 and 2.10).

midheightduring the 1982 Urahawa-ohi and 1995 Kobe (Priestley et al, 1995) earthquakes, as

Examples of the sudden collapse caused by shear failure, special efforts must be taken in new

a consequence of premature termination of the column longitudinal reinforcement. An

and retrofit designs to guard against it.

example is shown in Figure 2.11, where the flexural-shear failure apparent corresponds to the bar cutoff location at column mid height. Bar terminations was based on the deign moment envelop, without accounting for the effects of bars lap spliced at this location. The effects of rotational inertia increasing the moment at column mid height may also have been significant in this case. Failure of the 18 columns of the collapse section of the Hanshin expressway in the 1995 Kobe earthquake was also initiated by premature termination of 33% of the column longitudinal reinforcement at 20% of the column height. This forced the plastic hinge to from above the base, where it could not benefit from confinement provided by the strong footing, which was essential for survival of the columns, since very little confinement reinforcement was provided. This dramatic failure is shown in Figure 2.11.

Figure 2.9: Shear failure within plastic hinge region, 1971 San Fernando earthquake

15

16

Earthquakes of particular significance for their impact on bridge design include Anchorage (1964), Niigata (1964), Inangahua (1968), San Fernando (1971), Guatemala (1975), Fruili(1976), Edgecumbe (1987), Loma Prieta (1989), Philippines (1990), Costa Rica (1991), Hanshin-Awaji (1995). In the last decade (and particularly since the Loma Prieta earthquake in Northern California) researchers and practitioners has been able to improve state-of-art substantially and major code revisions have occurred, in such area as design philosophy, performance criteria, ground motion characterization, geotechnical design (site effects), inelastic analysis and capacity design procedures for concrete and steel structures. Whereas much of this activity has been directed towards the design of new bridges, there has also been significant progress in the evaluation and retrofit of existing structures. The first seismic provisions for bridges were formulated in 1926 after experiencing the destructive damage in the 1923 Kanto Earthquake. Since the first stipulations, the seismic regulations have been reviewed and amended many times. The Design Specifications of Steel Highway Bridges were first issued in 1939 and were revised in 1956 and 1964. Figure 2.11: Flexure-Shear failure at pier mid-height of Route 43/2 overpass, due to premature termination of longitudinal reinforcement; 1995 Kobe earthquake.

The seismic design related requirements were extremely limited in those days due to lack of scientific knowledge to earthquake engineering. Only the seismic lateral force of 20% of the

2.3 DEVELOPMENT OF SEISMIC CODES It has been said that the history of seismic design is also the history of damaging earthquakes. It is certainly true that after each major earthquake in which bridge and building structures are seriously damaged, the codes-of-practice change. There is a strong correlation throughout the world between the occurrence of major earthquakes and advancement in seismic design. In the early 1900’s when seismic effects were either not or poorly considered in design, the 1923 Kanto earthquake with a moment magnitude of 7.9 occurred in the Tokyo-Yokohama area (Kawashima, 2000). This earthquake caused large scale damage to buildings and infrastructure, where bridges collapsed due to tilting, overturning, and settlement of the foundations. Due to this earthquake, the importance of considering seismic effects in design was recognized for the first time (Kawashima, 2011). Each earthquake has tested the knowledge-of-the-day and where it has been found deficient; the lessons learned have led to

gravity force was included, and no other seismic design related provisions were presented in these Specifications. The 20% gravity force was used as a basic design force for long time. The first comprehensive seismic design provisions were issued in 1971 as the "Guide Specifications for Seismic Design of Highway Bridges." The 1964 Niigata Earthquake triggered development of the Guide Specifications. It was described in the Guide Specifications that the lateral force shall be determined depending on zone, importance and ground condition in the static lateral force method (seismic coefficient method) and structural response shall be further considered in the modified static lateral force method (modified seismic coefficient method). Evaluation of soil liquefaction was incorporated in view of the damage caused by the 1964 Niigata Earthquake. Several independent design methods that had been developed for substructures were first unified between 1964 and 1971 in the form of "Guide Specifications of Substructures." Consequently, seismic safety was considerably upgraded in the bridges.

improvements in the sate-of-the-art. But new discoveries and fresh insight have also come from the research community which in recent years has been particularly active at generating

The 1971 Guide Specifications of Substructures and the 1971 Guide Specifications

insight and understanding and communicating results to the practicing profession. Codes and

forSeismic Design was revised in 1980 in the form of “Part IV Substructures” and "Part V

design provisions seem to be under constant revise and particularly so in recent years.

Seismic Design" of "Design Specifications of Highway Bridges". The Part V was essentially the same with the 1971 Guide Specifications for Seismic Design, but an updated evaluation method for predicting soil liquefaction as well as a practical design method for foundations in

17

18

liquefying sands was included in Part V. The Design Specifications were revised in 1990.

REDARS (Werner et al., 2003) which can be conducted on large and small scale regions.

Various major revisions were included in the Part V Seismic Design reflecting the progress of

Highway transportation systems play a significant role in the overall impact that a seismic

bridge earthquake engineering. The first was a unification of the static lateral force method

event has on a region because it interconnects with the other infrastructure in the region. The

(seismic coefficient method) and the modified static lateral force method (modified seismic

various methods for vulnerability assessment that have been proposed in the past describe

coefficient method). This included the revision of the lateral force coefficient. The second

below.

was an introduction of check of strength and ductility capacities of reinforced concrete piers. Depending on the failure mechanism, strength and ductility capacities of reinforced concrete

2.4.1 EXPERT OPINION BASED ANALYSIS

piers were evaluated. This was the first requirement in Japan to check the nonlinear behavior

When the Applied Technology Council (ATC) developed the ATC-13 report, there was a

of bridges after yielding of structural members. Although this provision has not been

relatively small amount of recorded data available for use in the generation of damage

mandatory, this significantly contributed to the enhancement of the ductility of piers. The

probability matrices. The data shortage was present for various types of structures and

third was an introduction of the static frame method to evaluate the lateral force of multi-span

facilities including various types of lifeline components. This lack of information necessitated

continuous bridges. This enabled to consider three-dimensional behavior of a bridge in the

the use of expert opinion for the generation of these damage matrices. The ATC put together

equivalent static analysis. The fourth was the provisions of design response spectra for

a panel of 42 experts whom they could query concerning the various components of a typical

dynamic response analysis.

Californian infrastructure (ATC, 1985). Only four of the 42 experts were chosen to provide information for highway bridges. The questionnaires that were created queried the experts on

In seismic design of structures, it is important to have a clear vision on the seismic performance. The basic concept of design philosophy and seismic performance criteria is more or less similar among the codes, i.e., for small-to-moderate earthquakes bridges should be resisted within the elastic range of the structural components without significant damage, and bridges exposed to shaking from large earthquakes should not cause collapse (Buckle 1996).

the probability of a bridge being in one of seven damage states for a given Modified-Mercalli Intensity (MMI) value. They also asked the experts to rate themselves on their experience in the field using a scale from zero to ten. After the questionnaires were completed and analyzed, the results were given back to the experts for a second look. They were permitted to consider the overall results and compare them with their initial responses and make any modifications they felt were necessary. These results were then compiled and reported as the damage probability matrices (DPM) for bridges in the ATC-13 report and were subsequently

2.4 SEISMIC VULNERABILITY ASSESSMENT

used in the ATC- 25 report (ATC, 1991).

The seismic vulnerability of a structure can be described as its susceptibility to damage by

There are several major concerns with this methodology, one of which is the subjectivity of

ground shaking of a given intensity. The aim of a vulnerability assessment is to obtain the

the procedure. There is little, if any, correlation to actual earthquake damage reports and is

probability of a given level of damage to a given building type due to a scenario earthquake.

based solely on the experience and number of experts queried. Another concern is that the

Seismic risk describes the potential for damage or losses that a region is prone to experience

DPMs were created for only two classes of bridges, major (spans over 500 feet) and

following a seismic event. This is in contrast to seismic hazard, which quantifies the

conventional (spans under 500 feet). Thus, a high level of uncertainty is present but not

recurrence rates of different ground motions. Seismic risk can also be defined as the spatially

quantified in these results. This uncertainty comes as a consequence of human judgment and

and temporally integrated product of the seismic hazard, the value of assets and the fragility

also the coarseness of the bridge classes.

of assets (Jacob, 1992). Basoz and Kiremidjian (1996) present a seismic event time-line which illustrates the events that take place before and after a seismic event. The first of these events is to assess the seismic risk, which estimates the potential losses that may occur as a result of the remainder of the events on the time-line. The assessment of these potential losses is done through the use of seismic risk assessment tools such as HAZUS (FEMA, 2003) and

19

2.4.2 EMPERICAL METHOD Following the 1989 Loma Prieta and 1994 Northridge earthquakes, empirical bridge fragility curves became more common as a direct result of more complete ground motion and bridge damage data. Empirical fragility curves are generated from actual earthquake data. This

20

methodology has been presented and demonstrated by several groups of people for the Loma

2.4.3 ANALYTICAL METHOD

Prieta and Northridge earthquakes such as Basoz and Kiremidjian (1997), Der Kiureghian(2002), Shinozuka et al. (2003) and Elnashai et al. (2004) and by Shinozuka et al. (2000) and Yamazaki et al. (1999) for the 1995 Kobe earthquake. Although there are some slight variations in the methods used by the aforementioned groups they are conceptually the same. The procedure requires that a post-earthquake assessment be performed where a damage state would be assigned to all the bridges that belong to the bridge class being considered. A shake map, that geographically defines the ground motion in terms of some intensity measure, such as peak ground acceleration (PGA) is used to assign each bridge to a

When actual bridge damage and ground motion data are not available analytical fragility curves must be used to assess the performance of highway bridges. There have been many researchers that have developed analytical fragility curves for bridges using a variety of different methodologies. Since damage states are related to structural capacity (C) and the ground motion intensity parameter is related to structural demand (D), the fragility or probability of failure can then be described as in Equation 1. This specifically gives the probability that the seismic demand will exceed the structural capacity.

damage state and a given ground motion intensity in a damage frequency matrix. Table 3-1 is

|

|

(1)

an example of a damage frequency matrix that was assembled by Basoz and Kiremidjian (1997) for all multi-span bridges damaged during the 1994 Northridge earthquake. Table: 2.1 Damage Frequency Matrixes for Multi-Span Bridges (Basoz andKiremidjian, 1997).

This probability is generally modeled as a lognormal probability distribution. It is chosen because it has shown to be a good fit in the past and is convenient for manipulation using conventional probability theory (Wen et al., 2003). In addition, when the structural capacity and demand roughly fit a normal or lognormal distribution, using the central limit theorem, it can be said that the composite performance will be lognormally distributed (Kottegoda and Rosso, 1997). Thus the fragility curve can be represented by a lognormal cumulative distribution function which is given in Equation 2 (Melchers, 2001).

(

⁄ √

)

(2)

where S c is the median value of the structural capacity defined for the damage state, βc is the dispersion or lognormal standard deviation of the structural capacity, Sd is the seismic demand in terms of a chosen ground motion intensity parameter, βd is the logarithmic standard deviation for the demand and Φ [.] is the standard normal distribution function. Referring to Equation 2, it can be seen that the structural demand and capacity must be modeled to generate analytical fragility curves. There are a number of methodologies which have been used by researchers to accomplish this task. The methodologies they have employed range from simplistic to fairly rigorous. The following sections introduce some of Although this method is relatively straight forward it has some drawbacks and limitations.

these methodologies.

The first limitation is that it is difficult to get an adequate number bridges belonging to one bridge class that lie in a particular damage state. When this is the case it is difficult to get

2.5 RETROFIT STRATEGIES FOR BRIDGES

statistically significant results (Shinozuka, 1998). Thus, it is often required to group classes

Main purpose of the seismic retrofit of reinforced concrete columns is to increase their shear

together to get enough bridges in a given damage state and hence reduces the usefulness of

strength, in particular in the piers with termination of longitudinal reinforcements at the mid

the fragility curves.

21

22

height without enough development length. This enhances the ductility of columns because

is provided at the ends of the column to prevent the jacket from bearing on adjacent members.

premature shear failure could be avoided.

This serves to avoid undesirable flexural strength enhancement in which larger shears and moments may be transferred to the footings and cap beams under seismic loading (Priestley et

2.5.1 CONCRETE JACKETS

al., 1996). While the effect is not intended, experimental testing by Chai et al. has revealed

Concrete jacketing has been the method of choice for rehabitation of deficient structures. It is

that the steel jacket increases column sti_ness by approximately 10 to 15% for partial height

generally cheaper than other retrofit measures and it is also suitable method for retrofitting

(Chai et al., 1991) and 20 to 40% for full height jackets (Priestley et al., 1996). This could

columns in water. If applied with appropriate reinforcement, concrete jacket can enhance the

undesirably impact the impact force and performance of bridge components, and is thus a

stiffness, flexural and shear strength as well as the deformation capacity and are effective in

critical consideration for analytical assessment of this retrofit.

achieving enhance confinement. Concrete jacket has some disadvantages as compared of FRP and steel jacket as it increased the size of the structural members. Figure 2.12 illustrates the concrete jacketing technique.

Figure 2.13: Typical steel jacket retrofit details: (a) full height (b) typical section. The steel jackets are designed according to the Seismic Retrofitting Manual for Highway Figure 2.12: Concrete Jacket

Structures (FHWA, 1995), which considers the increase in compressive strength and ultimate strain in the concrete due to steel jacket confinement. The steel jackets are modeled by

2.5.2 STEEL JACKETS

altering the fiber section for the concrete column. In a fiber model, composite sections are Concrete columns may be encased in steel jackets to help overcome typical column deficiencies. Columns have been found to be particularly vulnerable due to insufficient lap splices and inadequate transverse reinforcement, leading to limited ductility capacity and low shear strength (DesRoches et al., 2000). Steel jacketing has been used as a retrofit measure to enhance the flexural ductility, shear strength, or performance of lap splices in reinforced

created with fibers representing the unique stress-strain relationships for the longitudinal steel reinforcement, unconfined concrete, and confined concrete. Thus, the concrete fibers now have enhanced compressive strength and ultimate strain due to the confinement caused by jacketing. This effect is illustrated in Figure 2.13, adapted from Priestley et al. (1996). The compressive strength of concrete, fꞌcc, is estimated from Chai et al. (1991) following

concrete bridge columns. Extensive proof of concept testing of steel jacketed bridge columns was performed at the University of California, San Diego in the early 1990s, and Priestley et

(

al. (1996) cite that several hundred bridges in the US had been retrofit with this technology by

√

)

(3)

1996. A review of the state-of-practice in the CSUS has revealed that this is the most common column retrofit in the CSUS, as well. Figure 2.1 details a typical cross section of a

Where,

is the radial confining stress in the steel jacket at yield given by,

circular column retrofit by a steel jacket, and the full height configuration which is assumed

(4)

for this study. The steel jackets are typically A36 steel casings and a space of about 50.8 mm

23

24

And

is the yield stress of the jacket, Dj is the jacket diameter, and tj is the jacket

thickness. In addition to the above modeling, the elastic modulus of the jacketed column section is increased such that there is a 20 to 40% increase in stiffness of the column based on Priestley’s (Priestley et al., 1994a) test results of jacketed columns. In general, while the analytical model is slightly affected by the use of steel jacketing, the primary impact of the retrofit is to increase the column ductility capacity.

Figure 2.15: Steel Jacket Retrofit at Expressway in 1989, which was effective during the 1995 Kobe earthquake Figure 2.14: Effect of confinement on concrete, adapted from Priestley et al. (1996).

2.5.3 PRECAST CONCRETE SEGMENT JACKET

The circular columns are retrofitted by circular steel jackets while rectangular columns are

Because steel jacket is vulnerable to corrosion, it is not generally used for retrofit of columns

retrofitted by elliptic or rectangular jackets, the former is more effective. In the case of

under water in river, lake and sea. As reinforced concrete jacket are used for retrofit of

circular columns, two half shells of steel plate are rolled to a radius of 12.5 to 25 mm bigger

columns in water, however it generally takes longer construction period. Setting a new

than the column radius (Priestley, 1996), and are site welded up the vertical seams. The steel

reinforced concrete jacket requires to dry up the top of footing and piers. Therefore a

shall provide a continuous tube around the concrete column with an annular gap. This gap is

reinforced concrete jacketing is costly. As a consequence a jacketing method which uses

grouted with cement grout or epoxy resin. A small gap of about 50 mm is provided at the

precast concrete segments is now increasingly used for columns in water. There are at least

bottom of piers between steel jacket and the top of the footing. This gap enables jacket to

three reasons for the wide acceptance of the precast concrete segment jacketing. First is the

function as a passive confinement and prevents excessive increase in the flexural strength.

technical development which enables to set precast concrete segments without drying up a

This jacket primarily increases shear strength and confinement. The increase of thickness of

foundation under water. Second is the faster construction than the standard reinforced

column after retrofit is very small.

concrete jacketing. Setting of prefabricated concrete segments contributes to significantly reduce the construction period. Special joints for connection of segments are sometimes used to further reduce construction period. Third is the cost saving. Because size of the columns is generally more or less the same at a bridge, fabrication of segments in a factory and setting them at the site save the cost compared to the reinforced concrete jacketing. In other word, the precast concrete segment is not competent at the bridges which are supported by irregular columns with different sections.

25

26

and pre-cured (composite sheets and shapes manufactured off-site). The properties of an FRP system shall be characterized as a composite, recognizing not just the material properties of the individual fibers, but also the efficiency of the fiber-resin system and fabric architecture. Table 1 indicates mechanical properties of the composites (FRP systems) in the direction of fibers formed by curing of unidirectional fibers having fiber-resin ratio of approximately 1:1 by volume and a thickness of 2.5mm. Table 2.2: Typical Mechanical Properties of FRP Laminates FRP System Aramid + Epoxy High performance Figure 2.16: Precast concrete segment jacket

Carbon + Epoxy High Strength E-Glass + Epoxy

Tensile Strength

Elastic Modulus

Ultimate Elongation

(MPa)

(GPa)

(%)

700-1725

48-68

2.0-3.0

1025-2075

100-145

1.0-1.5

525-1400

20-40

1.5-3.0

FRP strengthening is a quick, neat, effective, and aesthetically pleasing technique to rehabilitate reinforced and pre-stressed concrete structures. Unlike steel plates, FRP systems possess high strength to self-weight ratio and do not corrode. But, it is imperative to be aware of the performance characteristics of various FRP systems under different circumstances to select a durable and suitable system for a particular application. It should be ensured that the FRP system selected for structural strengthening has undergone durability testing consistent with the application environment.

Figure 2.17: Before retrofit

Figure 2.18: After retrofitted

2.5.4 FRP STRENGTHENING A Fiber Reinforced Polymer (FRP) typically consists of high tensile continuous fibers oriented in a desired direction in a specialty resin matrix. These continuous fibers are bonded to the external surface of the member to be strengthened in the direction of tensile force or as confining reinforcement normal to its axis. FRP can enhance shear, flexural, compression capacity and ductility of the deficient member. Aramid, carbon, and glass fibers are the most common types of fibers used in the majority of commercially available FRPs. FRP systems, commonly used for structural applications, come in many forms including wet lay-up (fiber sheets or fabrics saturated at site), pre-preg (pre-impregnated fiber sheets of fabrics off site)

27

28

deformation remains as residual strain upon unloading but can be removed upon heating. This is called shape memory effect. SMAs also possess super elastic characteristics. After pseudo yield the deformation can be recovered simply by unloading. This is called super elastic effect. The high elastic strain capabilities are of the order of 6-8%. The shape memory and super elastic effect depend on the temperature: In the lower range they show shape memory effect but in 20 to 300C higher than transformation temperature these materials show super elastic effect. The transformation temperature is a function of alloy compound. SMAs can act both as restrainers and dampers. SMAs can dissipate significant amount of energy due to shape memory effect (Andrawes and Desroches, 2004). One application of SMAs is in the use of restrainers. These devices are installed between adjoining spans to limit excessive relative displacements. Traditionally steel restraining cables have been used for this purpose. The disadvantages of traditional devices are, (i) these have no re-centering capability (ii) the devices increase ductility demand of structure. Both the drawbacks of traditional devices are Figure 2.19: FRP strengthening of concrete pier

overcome in restrainers made of SMA alloys.

2.5.5 EXTERNAL POST-TENSIONING Over the service life of a pre-stressed concrete member, loss in pres-stress may occur due to a variety of reasons. Post-tensioned bridges can be effectively rehabilitated by external posttensioning technique to compensate for loss in pre-stress or increase in wheel loads. In this technique, pre-stressing tendons or bars are located according to pre-determined profile on the external surface of the member to be strengthened according to design. Anchor heads are positioned at the ends of these tendons/ bars to post-tension the member using hydraulic jacks. Although, this method is quite effective but it requires sufficient strength in the existing concrete to transfer the stress and exposed tendons & anchorages need to be protected against corrosion and vandalism. Photo 10 shows external post tensioning for a bridge girder. 2.5.6 SMART MATERIALS The smart materials have unusual thermo mechanical properties that have been explored for purpose of earthquake protection and retrofitting measures in structures. These materials also called intelligent materials have self-repairable and self-diagnosis characteristics. Shape memory alloys (SMAs) are one kind of materials, which has found application in bridge structures. The SMA materials differ from ordinary materials in that in the former, heating the material can change the crystal structure. These materials can have phase transformation above a transformation temperature range that is specific to various alloys. One such material is use is Nitinol, which has shape memory effect and super elastic effect. Shape memory effect permits having pseudo yield and pseudo elastic deformation. The pseudo-elastic

29

30

CHAPTER 3 STRENGTH AND DUCTILITY OF BRIDGE BENT 3.1 GENERAL Bridges play very important role for evacuation and emergency routes for rescues, first aid, medical services, fire-fighting and transporting urgent disaster commodities. In view of importance of Highway Bridge in transportation network, it is the key issue to minimize as much as possible loss of the bridge functions during earthquakes. In the last few earthquakes, for instance, the Kobe earthquake in 1995, the Northridge earthquake in 1994, the Chi-Chi earthquake in 1999, and the Chile and Haiti earthquakes in 2010 have demonstrated that a number of highway bridges have collapsed or have been severely damaged, even though they were subjected to earthquake ground shaking of an intensity that has been frequently less than the current code intensities. Hence it is important to evaluate the strength and the performance of the bridge. In this chapter, the lateral strengths of pier are assessed. The lateral strengths of the piers are obtained from bending and the shear capacity. The bending strengths are obtained from pushover analysis results, while the shear strengths are estimated by using code defined equations. The capacity required for the current and future probable increase of spectrum acceleration is analyzed for safety evaluation of the bent. It was found from the analysis that the analyzed bent would suffer damage condition if seismic intensity exceeds current code practice.

3.2 DESCRIPTION OF THE HIGHWAY BRIDGE AND THE BENT Figure 3.1: 3D View of Bahadarhat Flyover

Chittagong city is surrounded by many primary and secondary road networks. In Chittagong Metropolitan Master Plan, there is a guideline for improvement of traffic network to reduce the

The piers heights blow the pile cap varies from 3.653m to7.29m. A geometric model of the

traffic congestion within the city. A 1331.60 meter long Flyover connecting CDA (Chittagong

flyover is presented in figure 3.2. The span length Varies from 35m to 40m. The diameter is 2.5m

Development Authority) Avenue road and Shah Amanat Bridge approach road was construct to

for all piers. The height, sectional dimensions, longitudinal reinforcement is presented in tabular

reduce traffic congestion of the junction. Figure 3.1 shows the three dimensional view of the

form in Table3.2.

Bahadarhat flyover. The very purpose of Flyover of Bahadarhat junction is to establish a quick and efficient movement of vehicles over the bridge.

31

32

Table 3.2: Material properties of the pier Material Name

Confined Concrete

Description of Properties Compressive strength = 35 MPa Tensile strength = 3 MPa Strain at Peak Stress = 0.0025 ( mm/mm) Confinement factor = 1.2 Specific weight = 2.4 E-05 N/mm3 Modulus of elasticity = 200000 MPa

Steel

Unconfined Concrete

Yield strength = 485 MPa Strain hardening parameter = 0.0075 Fracture/ buckling strain = 0.06 Specific weight = 7.8 E-05 N/mm3 Compressive strength = 35 MPa Tensile strength = 3 MPa Strain at Peak Stress = 0.0025 ( mm/mm) Confinement factor = 1.0 Specific weight = 2.4 E-05 N/mm3

3.3 ASSESSMENT OF BRIDGE BENT 3.3.1 PLASTIC HINGE LENGTH AND ULTIMATE DISPLACEMENT OF PIER To obtain the force-displacement relationship at the top of the bridge pier, the pier is divided into

Figure 3.2: Geometrical Model of Flyover

N slices (50 slices are recommended in the code) along its height. For sectional analysis, it is

Table 3.1: Geometries properties of the pier Pier No 1 2 3 4 5 to 14 21 23 24

Pier Height (m) 3.653 4.917 5.857 6.557 7.29 7.057 5.417 4.153

Diameter (m) 2.5 2.5 2.5 2.5 2.5 2.5 2.5 2.5

Long. Reinf. 72 @ D32 72 @ D32 72 @ D32 72 @ D32 72 @ D32 72 @ D32 72 @ D32 72 @ D32

mainly focused on three sections: 1) section at the top level, 2) section at one-third level from the bottom of the pier, and 3) section at the base level. This is because the configuration of the reinforcement at this level is different. Finally, the force displacement relationship at the top of the bridge pier is obtained using the moment-curvature diagrams and shear stress-strain diagram. Figure 3.4 shows the numerical evaluation of the flexural and shear components of displacement. The steps for obtaining the force-displacement relationships are as follows: 1. The pier is divided into N slices along its height. 2. The moment-curvature diagrams for each cross-section is obtained through sectional analysis.

The strengths and ductility of the piers largely depend on the material strengths and stress-strain relationship. Material strengths are found from the design data are presented in table 3.2.

33

3. The horizontal force P is applied at the top of the pier. 4. The bending moment diagrams of the pier for the applied force P is drawn.

34

5. The curvature from bending moment and moment-curvature diagram is obtained. 6. The displacement δ at the top of the pier is estimated using the following equation.

50000

7.In a similar way, several forces P are applied and the corresponding displacement obtain. Finally, using these values, the force –displacement relationship at pier top is obtained.

Moment (KNm)

40000

30000

20000

10000 Applied load on column

B.M.D Curvature diagram S.F.D

S.S.D

Figure 3.3: Numerical evaluation of flexural and shear component of displacement

0 0

0.005

0.01

0.015

0.02

0.025

0.03

In the current study a finite element based program Response-2000 has been used.

Curvature (rad/m) Evaluation of the adequacy of existing bridge bent to withstand imposed seismic loads requires assessment and comparison of anticipated demand and available capacities. The Lateral strengths of the piers in bending are obtained by using the ultimate moment capacities of the pier obtained from sectional analysis. The Moment curvature relationship shown below:

Figure 3.4: Moment-Curvature Relationship of Pier 1 The moment-curvature relationship of the pier 1 is illustrated in the figure 3.5. It is seen from Figures that the moment is found to increase rapidly with increasing curvatures initially. With a curvature of 0.005 rad/m the pier comes to its yield state. The yield moment of the pier is, My= 43000 KNm. Curvatures are observed to increase rapidly with little increase in moment after yielding. The ultimate moment and ultimate curvature are found from the graph are 47500 KNm and 0.028 rad/m respectively.

35

36

50000

50000

40000

30000

Moment (KNm)

Moment (KNm)

40000

20000

30000

20000

10000 10000 0 0

0.005

0.01

0.015

0.02

0.025

0.03

Curvature (rad/m)

0 0

0.01

0.02

0.03

0.04

Curvature (rad/m) Figure 3.5: Moment-Curvature Relationship of Pier 2 The moment-curvature relationship of the pier 2 is illustrated in the figure 3.6. It is seen from

Figure 3.6: Moment-Curvature Relationship of Pier 3

Figures that the moment is found to increase rapidly with increasing curvatures initially. With a curvature of 0.007 rad/m the pier comes to its yield state. The yield moment of the pier is, My= 41000 KNm. Curvatures are observed to increase rapidly with little increase in moment after yielding. The ultimate moment and ultimate curvature are found from the graph are 46200 KNm

The moment-curvature relationship of the pier 3 is illustrated in the figure 3.7. It is seen from Figures that the moment is found to increase rapidly with increasing curvatures initially. With a curvature of 0.008 rad/m the pier comes to its yield state. The yield moment of the pier is, My= 40000 KNm. Curvatures are observed to increase rapidly with little increase in moment after

and 0.028 rad/m respectively.

yielding. The ultimate moment and ultimate curvature are found from the graph are 45300 KNm and 0.030 rad/m respectively.

37

38

50000

50000

45000

40000

35000 Moment (KNm)

Moment (KNm)

40000

30000 25000 20000

15000

30000

20000

10000

10000

5000 0 0

0.01

0.02

0.03

0.04

Curvature (rad/m)

Figure 3.7: Moment-Curvature Relationship of Pier 4

0 0

0.01

0.02 Curvature (rad/m)

0.03

0.04

Figure 3.8 : Moment-Curvature Relationship of Pier 5-21

The moment-curvature relationship of the pier 4 is illustrated in the figure 3.8. It is seen from Figures that the moment is found to increase rapidly with increasing curvatures initially. With a curvature of 0.008 rad/m the pier comes to its yield state. The yield moment of the pier is, My= 40000 KNm. Curvatures are observed to increase rapidly with little increase in moment after yielding. The ultimate moment and ultimate curvature are found from the graph are 46200 KNm

The moment-curvature relationship of the pier 5-21 is illustrated in the figure 3.9. It is seen from Figures that the moment is found to increase rapidly with increasing curvatures initially. With a curvature of 0.009 rad/m the pier comes to its yield state. The yield moment of the pier is, My= 38000 KNm. Curvatures are observed to increase rapidly with little increase in moment after yielding. The ultimate moment and curvature are found from the graph are 44200 KNm and 0.037

and 0.034 rad/m respectively.

rad/m respectively.

39

40

60000

40000

50000 Moment (KNm)

Moment (KNm)

50000

30000

20000

40000 30000

20000

10000 10000

0 0

0.01

0.02

0.03

0.04

0 0

0.005

0.01

0.015

0.02

0.025

0.03

Curvature (rad/m) Curvature (rad/m) Figure 3.9 : Moment-Curvature Relationship of Pier 22 Figure 3.10: Moment-Curvature Relationship of Pier 23 The moment-curvature relationship of the pier 22 is illustrated in the figure 3.6. It is seen from Figures that the moment is found to increase rapidly with increasing curvatures initially. With a curvature of 0.006 rad/m the pier comes to its yield state. The yield moment of the pier is, My= 39000 KNm. Curvatures are observed to increase rapidly with little increase in moment after yielding. The ultimate moment and curvature are found from the graph are 45500 KNm and 0.032

The moment-curvature relationship of the pier 23 is illustrated in the figure 3.11. It is seen from Figures that the moment is found to increase rapidly with increasing curvatures initially. With a curvature of 0.005 rad/m the pier comes to its yield state. The yield moment of the pier is, My= 40500 KNm. Curvatures are observed to increase rapidly with little increase in moment after yielding. The ultimate moment and curvature are found from the graph are 46500 KNm and 0.028

rad/m respectively.

rad/m respectively.

41

42

a minor change in the slope is observed in the initial linear regime. It is due to developing tension cracks in the cover concrete, and hence reduction of effective cross-sectional area occurs. It is also seen that the trend of moment-curvature relationships are for the section same but the slopes,

60000

and the characteristic points are found different. The ultimate displacement

50000

of a pier is evaluated from yielding and ultimate curvature

and

Moment (KNm)

as (Priestley and Park 1987, Priestley at al. 1996)

40000 =

30000

+(

−

)(h −

/ 2)

(3.1)

In which h =height of the pier and

=plastic hinge length,

= 0.2h − 0.1D; (0.1D ≤Lp≤ 0.5D)

20000

(3.2)

Table 3.3: Plastic hinge length and ultimate displace of bent

10000 0 0

0.005

0.01

0.015

0.02

0.025

0.03

Curvature (rad/m)

Yield hinge Curvature, (m) (rad/m)

Ultimate Curvature

Ultimate displacement. (mm)

Pier No

Pier Height Plastic length, (m)

1

3.653

0.4806

0.005

0.028

45.72

((rad/m)

2

4.917

0.7334

0.007

0.028

82.08

Figure 3.11: Moment-Curvature Relationship of Pier 24

3

5.857

0.9214

0.008

0.03

127.39

The moment-curvature relationship of the pier 24 is illustrated in the figure 3.12. It is seen from

4

6.557

1.0614

0.008

0.034

196.30

Figures that the moment is found to increase rapidly with increasing curvatures initially. With a

5 to 21

7.29

1.208

0.009

0.037

269.15

curvature of 0.006 rad/m the pier comes to its yield state. The yield moment of the pier is, My=

22

7.057

1.1614

0.006

0.033

243.08

42000 KNm. Curvatures are observed to increase rapidly with little increase in moment after

23

5.417

0.8334

0.005

0.028

109.85

24

4.153

0.5806

0.006

0.028

58.34

yielding. The ultimate moment and curvature are found from the graph are 49500 KNm and 0.028 rad/m respectively. It is seen from above discussion that that the moment is found to increase rapidly before yielding

3.3.2 LATER STRENGTH AND DUCTILITY OF BRIDGE BENT

of the pier with increasing curvatures initially, while the rate of increase becomes insignificant after an interval. The reason for changing the relation is that reinforcing steel in the extreme tensile layer reaches yield strength. The moment in the stage is termed as yield moment. Moments are observed to increase further with curvature beyond the yield moment due to the fact that the reinforcement in layers other than in extreme layers is yet to reach yield strength. Further,

43

Evaluation of the adequacy of existing bridge bent to withstand imposed seismic loads requires assessment and comparison of anticipated demand and available capacities. With a view to achieve the goal, inelastic pushover analyses are carried out for obtaining the force-displacement relations. A force displacement relationship is generated by using seismostruct 2012. The

44

procedure illustrates in figure 3.13 to evaluated both yield and ultimate

displacement and

11000

capacity of the bent.

10000

9000 8000

Force (KN)

7000 6000 5000 4000 3000

2000 1000 0 0

10

20

30 40 Displacement (mm)

50

60

70

Figure 3.13: Force-displacement Relationship of Pier 01 The force-displacement relationship of the pier 1 is illustrated in the figure 3.13. It is seen from

(0,0)

δyδu

figure that the initially force is found to increase rapidly with increasing displacement. Pier comes to its yield state when displacement reaches 8mm. The yield force of the pier is, Fy= 8700 KN.

Displacement Figure 3.12: Pushover analysis of the bent

Displacement are observed to increase rapidly with little increase in force after yielding, at this state pier shows inelastic displacement. The ultimate Force is found from the graph are 9994.867 KN.

The pushover analysis results for the piers are presented in the form of load-displacement relationships that are presented in Figures below:

45

46

9000

7000

8000 7000

6000

6000 5000

4000

Force (KN)

Force (KN)

5000

3000 2000 1000

4000 3000 2000

0 0

50

100

150

200

250

-1000 Displacement (mm)

1000

0 0

20

60

80

100

120

140

Displacement (mm)

Figure 3.14: Force-displacement Relationship of Pier 2 The force-displacement relationship of the pier 2 is illustrated in the figure 3.14. It is seen from

40

Figure 3.15: Force-displacement Relationship of Pier 03

figure that the force is found to increase rapidly with increasing displacement initially before yielding of the pier. Pier comes to its yield state when displacement reaches 18mm. The yield

The force-displacement relationship of the pier 2 is illustrated in the figure 3.15. It is seen from

moment of the force is, Fy= 7200 KN. Displacement are observed to increase rapidly with little

Figure that the force is found to increase rapidly with increasing displacement initially. Pier

increase in force after yielding, at this state pier shows inelastic displacement. The ultimate Force

comes to its yield state when displacement reaches 20mm. The yield moment of the force is, Fy=

is found from the graph is 8136.065 KN.

6000 KN. Displacement are observed to increase rapidly with little increase in force after yielding, at this state pier shows inelastic displacement. The ultimate Force is found from the graph are 6328.135 KN.

47

48

5000

4500

4500

4000

4000

3500

3500

3000

3000 Force (KN)

Force (KN)

5000

2500 2000

2500 2000

1500

1500

1000

1000

500

500

0

0 0

50

100 150 Displacement (mm)

200

250

0

50

100

150

200

250

300

-500 Displacement (mm)

Figure 3.16: Force-displacement Relationship of Pier 04 The force-displacement relationship of the pier 04 is illustrated in the figure 3.16. It is seen from

Figure 3.17: Force-displacement Relationship of Pier 05-21

Figure that the force is found to increase rapidly with increasing displacement initially. Pier comes to its yield state when displacement reaches 8mm. The yield moment of the force is, Fy= 3900 KN. Displacement are observed to increase rapidly with little increase in force after yielding, at this state pier shows inelastic displacement. The ultimate force is found from the

The force-displacement relationship of the pier 05-21 is illustrated in the figure 3.14. It is seen from Figure that the force is found to increase rapidly with increasing displacement initially. Pier comes to its yield state when displacement reaches 8mm. The yield moment of the force is, Fy= 8500 KN. Displacement are observed to increase rapidly with little increase in force after

graph are 4358.049 KN.

yielding, at this state pier shows inelastic displacement. The ultimate Force is found from the graph are 4368.301 KN.

49

50

5000 4500

8000

4000 7000

3000

6000

2500

5000

Force (KN)

Force (KN)

3500

2000 1500

4000 3000

1000

2000

500

0

1000 0

50

100 150 Displacement (mm)

200

250

300

0 0

20

40

60

80

100

Displacement (mm)

Figure 3.18: Force-displacement Relationship of Pier 22 The force-displacement relationship of the pier 22 is illustrated in the figure 3.14. It is seen from

Figure 3.19: Force-displacement Relationship of Pier 23

Figure that the force is found to increase rapidly with increasing displacement initially. Pier comes to its yield state when displacement reaches 8mm. The yield force of the force is, Fy=

The force-displacement relationship of the pier 2 is illustrated in the figure 3.14. It is seen from

8500 KN. Displacement are observed to increase rapidly with little increase in force after

Figure that the force is found to increase rapidly with increasing displacement initially. Pier

yielding, at this state pier shows inelastic displacement. The ultimate force is found from the

comes to its yield state when displacement reaches 8mm. The yield force of the pier is, Fy= 8500

graph are 5976.935 KN.

KN. Displacement are observed to increase rapidly with little increase in force after yielding, at this state pier shows inelastic displacement. The ultimate force is found from the graph are 7945.387 KN.

51

52

Table 3.4: Lateral Strength of Bent 12000 Pier Height, h

Lateral strength, Pu (KN)

1

3.653

9994.867

2

4.917

8136.065

3

5.857

6328.135

4

6.557

4458.049

5 to 21

7.29

4368.301

22

7.057

5976.935

23

5.417

7945.387

24

4.153

9800.867

Pier No (m)

10000

Force (KN)

8000 6000 4000 2000

0 0

10

20

30

40

50

60

70 Ductility and lateral strength of the bent is evaluated from both Shear and flexural capacity of the

-2000

pier. Failure mode of bent is analyzed according to JRA, 2002. Strength and design displacement

Displacement (mm)

ductility factor

are determined depending on the failure mode as shown in Figure 3.21. Lateral

strength of the pier is analyzed and presented in the tabulated form in Table 3.3.

Figure 3.20: Force-displacement Relationship of Pier 24 The force-displacement relationship of the pier 24 is illustrated in the figure 3.14. It is seen from Figure that the force is found to increase rapidly with increasing displacement initially. Pier comes to its yield state when displacement reaches 8mm. The yield moment of the force is, Fy= 8500 KN. Displacement are observed to increase rapidly with little increase in force after yielding, at this state pier shows inelastic displacement. The ultimate force is found from the graph are 9800.867 KN and 0.028 rad/m respectively. Flexural strength of the piers is summarized from the above force-displacement analysis. In table 3.4 the lateral strength are presented. From the table it is evident that shorter piers possess greater flexural strength. The Pier lowest height has the greatest flexural capacity of 9994.86 KN whereas the highest pier with 7.29m height pier has only 4368.05 KN as lateral capacity.

53

54

Based on the flexural strength

Start

, shear strength Ps and shear strength under static loading

,

failure mode of a pier is decided to be one of the flexural failure, shear failure after flexural damage and shear failure as,

Compute ultimate flexural capacity,

{

(3.3)

{ The lateral capacity Pa and the design displacement ductility factor

Compute Shear capacity,

are given as

{

(3.4)

{

YES

(3.5)

NO In which α =safety factor depending on importance of bridges and the type of ground motion (α

Compute Shear capacity, assuming,

=3.0 and 2.4 for important and ordinary bridges, respectively, under the near field ground motions, and α =1.5 and 1.2 for important and ordinary bridges, respectively, under the Far field ground motions),

and

=yielding and ultimate displacement of the pier.

Shear strength of concrete can be calculated by following equation (JRA, 2002), (3.6)

YES Flexural failure

Shear failure After flexural cracks

NO (3.7)

Shear failure

(3.8) Where, = Shear Strength (N) = Shear Strength resisted by concrete (N) =Shear Strength borne by hoop tie (N)

End

a = Spacing of the stirrup (mm) The value of

and

given in Table 3.5,

Figure 3.21: Ductility Evaluation of Bent 55

56

Table 3.8: Failure mode: Shear failure of bridge pier Table 3.5: Value of

in relation to effective height d of a pier section (JRA, 2002) Pier height (m)

(KN)

1

3.653

9994.86

6699.565

2

4.917

8136.07

6699.565

23

5.417

7471.17

6489.565

24

4.153

9931.74

6699.565

Pier No Effective Height (mm)

Below 1000

3000

5000

Above 10000

1.0

0.7

0.6

0.5

(KN)

Failure Criteria

Failure mode Shear Failure Shear Failure

Table 3.6: Value of

in relation to effective height d of a pier section (JRA, 2002)

Shear Failure Shear Failure

Tensile Reinforcement (%)

0.2

0.3

0.5

Above 1%

0.9

1.0

1.2

1.5

Table 3.8 shows ductility of the bent that has been calculated by using figure 3.19. According to the steps shows in figure 3.19, bents with shear failure criteria will have allowable ductility 1.

The List of piers with flexural failure mode is summarized as Table 3.7. Piers taller than 6.5

The flexural ductility of the bent are calculated by using equation 3.7.

meter are vulnerable to flexural failure.

Table 3.9: Ductility of the Bent

Table 3.7: Failure mode: Flexural failure Pier No Pier Height (m)

(KN)

(KN)

Failure Criteria

Pier No

Pier (m)

Height

Failure mode

1

3.653

Shear Failure

1.0

2

4.917

Shear Failure

1.0

3

5.857

Flexural Failure

4.67

Failure mode

Failure Criteria

Allowable Ductility

3

6.557

6389.565 6289.565

Flexural Failure

4

6.557

4358.04

5985.565

Flexural Failure

5 to 21

7.29

4368.301 5985.565

Flexural Failure

4

6.557

Flexural Failure

4.56

22

7.057

4490.45

Flexural Failure

5 to 21

7.29

Flexural Failure

4.21

22

7.057

Flexural Failure

4.30

Table 3.8 demonstrates shear failure of the piers. Form the data it is evident that relatively short

23

5.417

Shear Failure

1.0

piers are more susceptible to shear failure rather than flexural failure.

24

4.153

Shear Failure

1.0

5985.565

57

58

The response spectrums of the earthquakes listed in Table 3.10 are shown in figure 3.20. Average

3.4 SAFETY EVALUATION OF THE BENT

response spectrum also shown in the figure. The highest response spectrum acceleration is For safety evaluation of the bent a suite of 20 far field ground motions are used in this study to develop response spectrum of this analysis. The characteristics of the earthquake ground motion

found to be 2.25 g and the average response spectrum has pick value just under 1.0g.

records are presented in Table 3.10. All these ground motions have very high PGA ranging from

2.5

0.2g to 0.7g. Figure 3.20 shows the acceleration response spectra of the recorded near fault

EQ1 EQ2 EQ3 EQ4 EQ5 EQ6 EQ7 EQ8 EQ9 EQ10 EQ11 EQ12 EQ13 EQ14 EQ15 EQ16 EQ17 EQ18 EQ19 EQ20 Average Average response spectrum

ground motions. Figure 3.21 shows the different percentiles of acceleration response spectra illustrating that the selected earthquake ground motion records are well describing the medium to

2

Table 3.10: Earthquake ground motion

Earthquake No EQ-1 EQ-2 EQ-3 EQ-4 EQ-5 EQ-6 EQ-7 EQ-8 EQ-9 EQ-10 EQ-11 EQ-12 EQ-13 EQ-14 EQ-15 EQ-16 EQ-17 EQ-18 EQ-19 EQ-20

Name

Northridge Landers Northridge Landers Duzce, Turkey Loma Prieta Hector Mine Loma Prieta Imperial Valley Manjil, Iran Imperial Valley Superstition Hills Kobe, Japan Superstition Hills Kobe, Japan Cape Mendocino Kocaeli, Turkey Chi-Chi, Taiwan Kocaeli, Turkey Chi-Chi, Taiwan

Recording Station

Beverly Hills - Mulhol Yermo Fire Station Canyon Country-WLC Coolwater Bolu Capitola Hector Gilroy Array #3 Delta Abbar El Centro Array #11 El Centro Imp. Co. Nishi-Akashi Poe Road (temp) Shin-Osaka Rio Dell Overpass Duzce CHY101 Arcelik TCU045

PGAmax (g)

PGVmax (cm/s.)

0.416 0.24 0.4 0.283 0.7 0.53 0.3 0.56 0.2 0.51 0.4 0.36 0.5 0.45 0.2 0.385 0.3 0.353 0.2 0.474

58.95 51.5 43.0 26 56.4 35 28.6 36 26.0 43 34.4 46.4 37.3 35.8 38.0 43.8 59.0 70.65 17.7 36.7

Spectral Acceleration(g)

strong intensity earthquake motion histories.

1.5

1

0.5

0 0.00

0.50

1.00

1.50

2.00 2.50 Period (sec)

3.00

3.50

Figure 3.22: Spectral Acceleration Figure 3.21 compares with different percentile of far field earthquake represented in Table 3.10 and BNBC response spectrum with different zone factor. A percentile of spectrum acceleration is a measure used in statistics indicating the value below which a given percentage of observations in a group of observations fall. The 25th percentile is the value below which 25 percent of the observations may be found. The 25th percentile is also known as the first quartile, the 50th percentile as the median or second quartile, and the 75th percentile as the third quartile.

59

60

4.00

Table 3.11: Safety Evaluation of piers for, Z=0.15 1.6

average

Spectral Acceleration, (m/s2)

equivalent response acceleration,

Acceleration Response Spectra(g)

75th BNBC, Z=.15 BNBC, Z=.28(Proposed) BNBC, Z=.36(Proposed)

0.8

Lateral Capacity, (KN)

Safety Status

Pier No

Allowable Ductility

1 2 3

1.0 1.0 1.0

1 1 2.87

3.68 3.68 3.68

3.68 3.68 1.28

(KN) 5625 7500 3895.556

9994.867 8136.065 6328.135

Safe Safe Safe

4

4.56

2.85

3.68

1.29

3947.368

4358.049

Safe

5 to 4.21 21

2.72

3.68

1.35

4136.029

4368.301

Safe

22 23 24

2.76 1 1

3.68 3.68 3.68

1.33 3.68 3.68

4076.087 7500 5625

5976.935 7945.387 9994.867

Safe Safe Safe

25th

1.2

Lateral Force Demand ( )

Response modification factor, R

4.30 1.0 1.0

Table 3.10 represents safety status of the piers for zonal factor, Z=0.28. This factor was taken into

0.4

consideration for the future probable change for seismic design for Chittagong. From the Table it can be seen that all the piers would be in “Not Safe” stage.

Table 3.12: Safety Evaluation of piers for, Z=0.28 0 0.00

0.50

1.00

1.50

2.00

2.50

3.00

3.50

4.00

Spectral Acceleration, (m/s2)

equivalent response acceleration,

Lateral Force Demand ( )

Lateral Capacity, (KN)

Safety Status

Pier No

Allowable Ductility

Response modification factor, R

Figure 3.23: Spectral Acceleration (percentile and BNBC response spectrum)

1 2 3

1.0 1.0 1.0

1 1 2.87

6.867 6.867 6.867

6.867 6.867 2.377848

(KN) 10500 14000 7271.705

9994.867 8136.065 6328.135

Not Safe Not Safe Not Safe

Table 3.9 represents safety status of the piers. The allowable ductility of the piers is found from

4

4.56

2.85

6.867

2.409474

7368.421

4358.049

Not Safe

5 to 4.21 21

2.72

6.867

2.524632

7720.588

4368.301

Not Safe

22 23 24

2.76 1 1

6.867 6.867 6.867

2.488043 6.867 6.867

7608.696 14000 10500

5976.935 7945.387 9994.867

Not Safe Not Safe Not Safe

Period(sec)

Table 3.9 and response spectrum of BNBC, Z=0.15g is used. From the Table it can be seen that all the piers are in safe status.

61

4.30 1.0 1.0

62

Table 3.10 represents safety status of the piers for zonal factor, Z=0.36. This factor was taken into consideration for the future probable maximum change for seismic zonal factor in Bangladesh.

CHAPTER 4 SEISMIC PERFORMANCE ANALYSIS OF THE BENT

From the Table it can be seen that all the piers would be in “Not Safe” stage.

4.1 GENERAL Table 3.13: Safety Evaluation of piers for, Z=0.38

In the last few decades, a dramatic increase in the losses caused by natural catastrophes has been observed worldwide. Reasons for the increased losses are manifold, though these certainly

Pier No

Allowable Ductility

Response modification factor, R

Spectral Acceleration, (m/s2)

equivalent response acceleration,

Lateral Force Demand ( )

include the increase in world population, the development of new “super-cities” , many of which Lateral Capacity, (KN)

Safety Status

(KN)

are located in zones of high seismic hazard, and the high vulnerability of modern societies and technologies. The 1994 Northridge (California, US) earthquake produced the highest ever insured earthquake loss and 1995 Kobe (Japan) earthquake was the highest ever absolute earthquake loss

1

1.0

1

9.3195

9.3195

14250

9994.867

2

1.0

1

9.3195

9.3195

19000

8136.065

3

1.0

2.87

9.3195

3.227079

9868.743

6328.135

4

4.56

2.85

9.3195

3.27

10000

4358.049

5 to 21

4.21

2.72

9.3195

3.426287

10477.94

4368.301

22

4.30

2.76

9.3195

3.37663

10326.09

5976.935

23

1.0

1

9.3195

9.3195

19000

7945.387

24

1.0

1

9.3195

9.3195

14250

9994.867

Not Safe Not Safe Not Safe Not Safe Not Safe Not Safe Not Safe Not Safe

(Calvi et al., 2006). Seismic performance estimating in future earthquakes are of fundamental importance for emergency planners and retrofitting of structures. With the occurrence of every major earthquake an increase social awareness of society’s seismic vulnerability is witnessed. The numerical results have revealed that the seismic responses of the bridge system are significantly affected by the characteristics of the earthquakes ground motion records (Bhuiyan at el., 2013). As a result seismic performance has become much research interest to quantify the potential social and economic losses of communities across the nation. Performance evaluation of bridge has grown from this surge in research as they are an essential component to the risk assessment methodology. This is done by estimating the performance of the various highway bridges in the network as a function of a ground motion intensity parameter. In this study twenty far field ground motions are used to evaluate bridge seismic performance.

3.5 SUMMARY The aim of this chapter was to evaluate the seismic capacity of a bridge subjected to current and

4.2 MODELING OF THE BRIDGE BENT

proposed future code measures and the spectral acceleration is compared with 20 far field

4.2.1 PHYSICAL MODEL OF THE BENT

earthquakes ranging from 0.2g to 0.728g. From result obtained from the analysis the capacity of the bridge was verified. Based on the results that were obtained the following conclusions were deduced:

To evaluate the seismic performance of the bridge bent, Bahadarhat flyover considered in this study. A typical 40m span with 7.29m high pier is considered for the study. The bent’s geometric configuration is shown in figure 4.1. Seven girders and a concrete deck are spanning between the

1. The bridge piers below 5.217m are susceptible to shear failure whereas piers more than

two bents.

5.217 are more likely to fail by flexure. 2. The lateral capacity of the pier is capable of withstand earthquake described by current code practice, that is Z=0.15g.

3. The lateral capacity of the pier will not be capable to withstand earthquake for the spectral acceleration of Z=0.28g and Z= 0.36g.

63

64

Here, fiber modeling approach has been employed to represent the distribution of material nonlinearity along the length and cross-sectional area of the member. The confinement effect of the concrete section is considered on the basis of reinforcement detailing. To develop the analytical model Menegotto-Pinto steel model (Menegotto and Pinto, 1973) with Filippou (Filippou et al., 1983) isotropic strain hardening property is used for reinforcing steel material.

Figure 4.1: Geometry of Span of the Flyover 4.2.2 ANALYTICAL MODEL OF THE BENT The analytical model of a bridge bent along with a bridge pier is shown in Figure 4.2. The analytical model of the bridge bent is approximated as a continuous 2-D finite element frame using the SeismoStruct nonlinear analysis program (SeismoStruct, 2010). 2-D inelastic beam elements have been used for modeling the bridge component. This simplification holds true only when the bridge superstructure is assumed to be rigid in its own plane which shows no significant structural effects on the seismic performance of the bridge system when subjected to earthquake ground acceleration in longitudinal direction (Ghobarah et at, 1988).

Figure 4.3: Menegotto-Pinto steel model FRP confined concrete model developed by Ferracuti and Savoia (2005) has been implemented. In this model the confinement effect of the FRP wrapping follows the rules proposed by Spoelstra and Monti (1999). The pier is modeled by using beam column element. Fiber Model is used to generate the section of the pier. In the current study, the nonlinear Fiber section is used to model the concrete pier.

Figure 4.2: Analytical Model of Bent 65

Figure 4.4: Fiber Section of Reinforced Pier 66

Table 4.1: Characteristics of far field ground motion histories.

Figure 4.5: Beam element for pier modeling

The pier and pier cap are modeled by using beam element. Pier cap is considered as solid concrete element for simplicity. Fiber section is used to model the concrete pier cap. The following figure shows the modeling of pier cap.

EQ No

Earthquake Name

1

Northridge

2

Landers

3

Northridge

4

Landers

5

M

Epicentral Distance (km)

PGAmax (g)

PGVmax (cm/s.)

6.7

13.3

0.416

58.95

7.3

86

0.24

51.5

6.7

26.5

0.41

42.97

Coolwater

7.3

82.1

0.283

26

Duzce, Turkey

Bolu

7.1

41.3

0.728

56.44

6

Loma Prieta

Capitola

6.9

9.8

0.53

35

7

Hector Mine

Hector

7.1

26.5

0.266

28.56

8

Loma Prieta

Gilroy Array #3

6.5

31.4

0.56

36

9

Imperial Valley

Delta

7.4

33.7

0.238

26

10

Manjil, Iran

Abbar

6.5

40.4

0.51

43

6.5

29.4

0.364

34.44

6.9

35.8

0.36

46.4

Twenty Far field ground motions are used in the analysis listed in Table 4.1. The ranges of epicentral distance are between 13km to 98km and the moment magnitude, Mw= 6.5-7.6. The peak ground acceleration (PGA) of the ground motions ranges from PGA 0.22 to PGA 0.728. In figure 4.1, Spectral acceleration with 5% damping is shown.

67

Beverly Hills Mulhol Yermo Fire Station Canyon Country-WLC

El Centro Array #11 El Centro Imp. Co.

11

Imperial Valley

12

Superstition Hills

13

Kobe, Japan

Nishi-Akashi

7.5

8.7

0.51

37.28

14

Superstition Hills

Poe Road (temp)

7.5

11.2

0.45

35.8

15

Kobe, Japan

Shin-Osaka

7.6

46

0.24

38

16

Cape Mendocino

Rio Dell Overpass

6.7

22.7

0.385

43.8

17

Kocaeli, Turkey

Duzce

7.3

98.2

0.312

59

18

Chi-Chi, Taiwan

CHY101

6.7

32

0.353

70.65

19

Kocaeli, Turkey

Arcelik

7.3

53.7

0.22

17.69

20

Chi-Chi, Taiwan

TCU045

7.1

77.5

0.474

36.7

Figure 4.6: Fiber Section of Reinforced Pier cap 4.3 EARTHQUAKE GROUND MOTION

Recording Station

68

4.4 TIME HISTORY ANALYSIS OF BENT Dynamic analysis is commonly used to predict the nonlinear inelastic response of a structure

Displacement (mm)

subjected to earthquake loading. The direct integration of the equations of motions is accomplished using the numerically dissipative α- integration algorithm (Hilber et al., 1977) or a special case of the former, the well-known Newmark scheme (Newmark,1959), with automatic time-step adjustment for optimum accuracy and efficiency. Modeling of seismic action is achieved by introducing acceleration loading curves (accelerograms) at the supports, noting that different curves can be introduced at each support, thus allowing for representation of asynchronous ground excitation.

Modeling of seismic action is achieved by introducing

acceleration loading curves (accelerograms) at the supports, noting that different curves can be

250 200 150 100 50 0 -50 -100 -150 -200 0

introduced at each support, thus allowing for representation of asynchronous ground excitation.

5

10 15 Time (Second)

20

25

30

Twenty different time histories earthquake tabulated in Table 4.1 are considered in the subsequent

Figure 4.8: Pier displacement history due to EQ-1

analysis to evaluate seismic performance. Time history analysis of the bent is conduct with the help of seismostruct 2012 software. The pier displacement diagrams are presented in the figures

Figure 4.7 represent acceleration history of Northridge earthquake from station location Beverly

below:

Hills with 13.3 epicentral distance. The maximum PGA of the earthquake is 0.416g and maximum PGV is 58.95 cm/sec. From the time history analysis it can be seen from the figure 4.8 that the bent have a maximum displacement of 200mm.

0.5

0.4 0.2 0.1 0

-0.1 -0.2 -0.3 -0.4 0

5

10

15 Time (Second)

20

Figure 4.7: Time history graph of EQ-1

25

30

Acceleration (g)

Acceleration (g)

0.3 0.2 0.15 0.1 0.05 0 -0.05 -0.1 -0.15 -0.2 -0.25 -0.3 0

10

20 Time (Second)

Figure 4.9: Time history graph of EQ-2 69

70

30

40

150

80 Displacement (mm)

Displacement (mm)

100 50 0

30 -20 -70

-50 -120

-100 0

10

20 Time (Second)

30

0

40

10 Time (Second)

15

20

Figure 4.12: Pier displacement due to EQ-3

Figure 4.10: Pier displacement due to EQ-2 Figure 4.9 represent Landers earthquake acceleration history record from Yermo Fire Station which has 86 km epicentral distance. The maximum PGA of the recorded earthquake is 0.24g and maximum PGV is 51.5 cm/sec. From the time history analysis it can be seen from the figure 4.10 that the bent have a maximum displacement of 105mm.

Figure 4.12 represent Duzce, Turkey earthquake acceleration history record from Bolu station which has 41.3 km epicentral distances. The maximum PGA of the recorded earthquake is 0.41g and maximum PGV is 42.47 cm/sec. From the time history analysis it can be seen from the figure 4.13 that the bent have a maximum displacement of 120 mm.

0.3

0.5 0.4 0.3 0.2 0.1 0 -0.1 -0.2 -0.3 -0.4

0.2 Acceleration (g)

Acceleration (g)

5

0.1 0 -0.1 -0.2 -0.3 0

0

5

10 Time (Second)

15

20

10

20 Time (Second)

Figure 4.11: Time history graph of EQ-3

Figure 4.13: Time history graph of EQ-4

71

72

30

40

150 80 100

Displacement (mm)

Displacement (mm)

60 40 20 0 -20 -40 -60 -80 0

10

20 Time (Second)

30

50 0 -50 -100 -150

40

0

5

10 15 Time (Second)

20

25

Figure 4.14: Pier displacement due to EQ-4

Figure 4.16: Pier displacement due to EQ-5

Figure 4.14 represent Landers earthquake time history record from Coolwater station which has

Figure 4.18 represent Duzce, Turkey earthquake time history record from Bolu station which has

82.1 km epicentral distances. The maximum PGA of the recorded earthquake is 0.283g and

41.3 km epicentral distances. The maximum PGA of the recorded earthquake is 0.728g and

maximum PGV is 26 cm/sec. From the time history analysis it can be seen from the figure 4.15

maximum PGV is 56.44 cm/sec. From the time history analysis it can be seen from the figure

that the bent have a maximum displacement of 72 mm. Due to great reduction in velocity it has

4.19 that the bent have a maximum displacement of 140 mm.

lower affect in bent displacement than the same earthquake recorded from Yermo Fire station.

0.8

0.5 0.3

0.4 Acceleration (g)

Acceleration (g)

0.6

0.2 0 -0.2 -0.4

0.1 -0.1 -0.3 -0.5

-0.6 0

5

10 15 Time (Second)

20

25

-0.7 0

5

10 15 Time (Second)

Figure 4.15: Time history of EQ-5

Figure 4.17: Time history of EQ-6

73

74

20

25

80

170

60

Displacement (mm)

Displacement (mm)

120 70 20 -30 -80

40 20 0 -20 -40 -60

-130

-80 0

5

10 15 Time (Second)

20

25

0

10 15 Time (Second)

20

25

Figure 4.20: Pier displacement due to EQ-7

Figure 4.18: Pier displacement due to EQ-6

Figure 4.22 represent Hector Mine earthquake from station location Hector from 26.5 km

Figure 4.20 represent Loma Prieta earthquake time history record from Capitola station which has

epicentral distances. The maximum PGA of the earthquake is 0.266g and maximum PGV is 28.56

9.8 km epicentral distances. The maximum PGA of the recorded earthquake is 0.53g and

cm/sec. From the time history analysis can is seen from the figure 4.23 that the bent have a

maximum PGV is 35 cm/sec. From the time history analysis it can be seen from the figure 4.21

maximum displacement of 75mm.

that the bent have a maximum displacement of 120 mm.

0.6

0.3

0.4 Acceleration (g)

0.2 Acceleration (g)

5

0.1 0 -0.1

0.2 0

-0.2 -0.4

-0.2 -0.6 -0.3 0

5

10

15

Time (Second)

20

25

-0.8 0

5

10 Time (Second)

15

Figure 4.19: Time history of EQ-7 Figure 4.21: Time history of EQ-8 75

76

20

25

Displacement (mm)

Displacement (mm)

120 100 80 60 40 20 0 -20 -40 -60 -80 0

5

10 15 Time (Second)

20

100 80 60 40 20 0 -20 -40 -60 -80 -100 -120 0

25

Figure 4.24 represent Loma Prieta earthquake from station location Gilroy Array #3 from 31.4 km epicentral distances. The maximum PGA of the earthquake is 0.56g and maximum PGV is 36 cm/sec. From the time history analysis can is seen from the figure 4.25 that the bent have a

50

Figure 4.26 represent Imperial Valley earthquake from station location Delta from 33.7 epicentral distances. The maximum PGA of the earthquake is 0.238g and maximum PGV is 26 cm/sec. From the time history analysis can is seen from the figure 4.27 that the bent have a maximum

Acceleration (g)

0.3

0.2 Acceleration (g)

40

displacement of 100mm.

maximum displacement of 100mm.

0.1 0 -0.1 -0.2

0.5 0.4 0.3 0.2 0.1 0 -0.1 -0.2 -0.3 -0.4 -0.5

0

-0.3 10

20 30 Time (Second)

Figure 4.24: Pier displacement due to EQ-9

Figure 4.22: Pier displacement due to EQ-8

0

10

20 30 Time (Second)

40

50

5

10 Time (Second)

Figure 4.25: Time history of EQ-10

Figure 4.23: Time history of EQ-9 77

15

78

20

25

30

Displacement (mm)

Displacement (mm)

80

-20 -70 -120 0

5

10 15 Time (Second)

20

25

100 80 60 40 20 0 -20 -40 -60 -80 -100

0

Time (Second) 10

5

Figure 4.26: Pier displacement due to EQ-10

15

20

25

Figure 4.28: Pier displacement due to EQ-11

Figure 4.28 represent Manjil, Iran earthquake from station location Abbar from 40.4 epicentral Figure 4.30 represent Imperial Valley earthquake from station location El Centro Array #11 from

distances. The maximum PGA of the earthquake is 0.51g and maximum PGV is 43 cm/sec. From

29.4 epicentral distances. The maximum PGA of the earthquake is 0.364g and maximum PGV is

the time history analysis can is seen from the figure 4.29 that the bent have a maximum

34.44 cm/sec. From the time history analysis can is seen from the figure 4.31 that the bent have a

displacement of 115mm.

maximum displacement of 88mm.

0.4 0.4

0.3

0.2

0.1

Acceleration (g)

Acceleration (g)

0.3 0.2

0

-0.1 -0.2

0.1 0 -0.1 -0.2 -0.3

-0.3

-0.4

Time (Second)

-0.4 0

5

10

15

20

25

0

5

10 15 Time (Second)

Figure 4.27: Time history of EQ-11 Figure 4.29: Time history of EQ-12 79

80

20

200

100

Displacement (mm)

Displacement (mm)

150

50

0 -50 -100 0

5

10 Time (Second)

15

20

140 120 100 80 60 40 20 0 -20 -40 -60 0

5

10

15

20

25

Time (Second)

Figure 4.30: Pier displacement due to EQ-12 Figure 4.32 represent Superstition Hills earthquake from station location El Centro Imp. Co. from

Figure 4.32: Pier displacement due to EQ-13

35.8 epicentral distances. The maximum PGA of the earthquake is 0.36g and maximum PGV is

Figure 4.34 represent Kobe, Japan earthquake from station location Nishi-Akashi from 8.7 km

46 cm/sec. From the time history analysis can is seen from the figure 4.33 that the bent have a

epicentral distances. The maximum PGA of the earthquake is 0.51g and maximum PGV is 37.28

maximum displacement of 150mm.

cm/sec. From the time history analysis can is seen from the figure 4.35 that the bent have a maximum displacement of 130mm.

0.4 0.3 0.4 0.3

0.1

0.2

0 -0.1 -0.2 -0.3 -0.4 0

5

10 15 Time (Second)

20

25

Acceleration (g)

Acceleration (g)

0.2

0.1 0 -0.1 -0.2 -0.3 -0.4 0

5

10 15 Time (Second)

Figure 4.31: Time history of EQ-13 Figure 4.33: Time history of EQ-14 81

82

20

25

80 60 40 Displacement (mm)

Displacement (mm)

100 80 60 40 20 0 -20 -40 -60 -80 -100

20 0 -20 -40 -60 -80

0

5

10 15 Time (Second)

20

-100

25

0

5

10

15

20 25 Time (Second)

30

35

40

Figure 4.34: Pier displacement due to EQ-14

Figure 4.36: Pier response due to EQ-15

Figure 4.36 represent Superstition Hills earthquake from station location Poe Road (temp) from

Figure 4.38 represent Kobe, Japan earthquake from station location Shin-Osaka from 46 km

11.2 km epicentral distances. The maximum PGA of the earthquake is 0.45g and maximum PGV

epicentral distances. The maximum PGA of the earthquake is 0.24g and maximum PGV is 38

is 36 cm/sec. From the time history analysis can is seen from the figure 4.37 that the bent have a

cm/sec. From the time history analysis can is seen from the figure 4.39 that the bent have a

maximum displacement of 90mm.

maximum displacement of 95mm.

0.3 0.2

0.1

0.1 Acceleration (g)

Acceleration (g)

0.3 0.2

0 -0.1 -0.2

0 -0.1 -0.2 -0.3 -0.4

-0.3 0

10

20 Time (Second)

30

40

-0.5 0

5

10

15

20

Time (Second) Figure 4.35: Time history of EQ-15 Figure 4.37: Time history of EQ-16 83

84

25

30

35

80 60 Displacement (mm)

Displacement (mm)

140 120 100 80 60 40 20 0 -20 -40 -60 -80

40 20 0 -20 -40

-60 0

5

10

15 20 Time (Second)

25

30

0

35

5

Figure 4.38: Pier response due to EQ-16

20

25

Figure 4.40: Pier response due to EQ-17

Figure 4.40 represent Cape Mendocino earthquake from station location Rio Dell Overpass from 22.7 km epicentral distances. The maximum PGA of the earthquake is 0.385g and maximum PGV is 44 cm/sec. From the time history analysis can is seen from the figure 4.41 that the bent have a maximum displacement of 120mm.

Figure 4.42 represent Kocaeli, Turkey earthquake from station location Duzce from 98.2 km epicentral distances. The maximum PGA of the earthquake is 0.312g and maximum PGV is 59 cm/sec. From the time history analysis can is seen from the figure 4.43 that the bent have a maximum displacement of 78mm.

0.3

0.8

0.2

0.6

0.1

0.4

0

Acceleration (g)

Acceleration (g)

10 15 Time (Second)

-0.1 -0.2

-0.3 0

5

10 15 Time (Second)

20

25

0.2 0

-0.2 -0.4 -0.6 0

Figure 4.39: Time history of EQ-17

5

10 15 Time (Second)

Figure 4.41: Time history of EQ-18 85

86

20

25

Displacement (mm)

Displacement (mm)

80 60 40 20 0 -20 -40 -60 -80 -100 0

5

10 15 Time (Second)

20

25

25 20 15 10 5 0 -5 -10 -15 -20 -25 0

Figure 4.42: Pier response due to EQ-18

5

10 15 Time (Second)

20

25

Figure 4.44: Pier response due to EQ-19

Figure 4.44 represent Chi-Chi, Taiwan from station location CHY101 from 32km epicentral distances. The maximum PGA of the earthquake is 0.535g and maximum PGV is 71 cm/sec. From the time history analysis can is seen from the figure 4.45 that the bent have a maximum displacement of 90mm.

Figure 4.46 represent Kocaeli, Turkey from station location Arcelik from 53.7 km epicentral distances. The maximum PGA of the earthquake is 0.22g and maximum PGV is 17.69 cm/sec. From the time history analysis can is seen from the figure 4.47 that the bent have a maximum displacement of 22mm.

0.4 0.2

0.3 0.2

Acceleration (g)

Acceleration (g)

0.1 0 -0.1

0.1

0 -0.1 -0.2 -0.3

-0.2

-0.4 0

-0.3 0

5

10 15 Time (Second)

20

5

10 15 Time (Second)

25 Figure 4.45: Time history of EQ-20

Figure 4.43: Time history of EQ-19 87

88

20

25

Displacement (mm)

100

Maximum displacements of the piers are obtained from the time history analysis. Displacement

50

ductility of the bent is calculated by using equation 4.1 and the calculated displacement compared

0

with Hwang et al. damage state. The results obtained are tabulated are in Table 4.4 shown below:

-50 TABLE 4.3: Damage state of bridge pier

-100 -150

-200 0

5

10 Time (Second)

15

20

25

Figure 4.46: Pier response due to EQ-20 Figure 4.48 represent Chi-Chi, Taiwan earthquake from station location TCU045 from 77.5 km epicentral distances. The maximum PGA of the earthquake is 0.474g and maximum PGV is 37 cm/sec. From the time history analysis can is seen from the figure 4.49 that the bent have a maximum displacement of 150mm.

4.5 ASSESSMENT OF BENT DAMAGE STATE The displacement ductility factor,

of the target bridge bent when subjected to the 20 mentioned

ground motions are evaluated to verify the seismic performance of the bridge structure. The displacement ductility factors are calculated as follows: (4.1) To analyze the damage state of the bridge bent damage state given by the Hwang et al. 2001 is considered as tabulated in Table 4.3.

TABLE 4.2: Damage/limit state of bridge components (Hwang et al, 2001) Damage State

Slight (DS=1)

Moderate (DS=2)

Extensive (DS=3)

Damage ((DS=3)

Cracking and Spalling

Moderate cracking and spalling

Degradation without collapse

Failure leading to collapse

Bridge Component

Physical Phenomenon

Earthquake No

Earthquake Name

Recording Station

PGAmax (g)

Disp. Ductility

Damage State

EQ-1 EQ-2 EQ-3 EQ-4 EQ-5 EQ-6 EQ-7 EQ-8 EQ-9 EQ-10 EQ-11 EQ-12 EQ-13 EQ-14 EQ-15 EQ-16 EQ-17 EQ-18

Northridge Landers Northridge Landers Duzce, Turkey Loma Prieta Hector Mine Loma Prieta Imperial Valley Manjil, Iran Imperial Valley Superstition Hills Kobe, Japan Superstition Hills Kobe, Japan Cape Mendocino Kocaeli, Turkey Chi-Chi, Taiwan

Beverly Hills - Mulhol Yermo Fire Station Canyon Country-WLC Coolwater Bolu Capitola Hector Gilroy Array #3 Delta Abbar El Centro Array #11 El Centro Imp. Co. Nishi-Akashi Poe Road (temp) Shin-Osaka Rio Dell Overpass Duzce CHY101

0.416 0.24 0.41 0.283 0.728 0.53 0.266 0.56 0.238 0.51 0.364 0.36 0.51 0.45 0.24 0.385 0.312 0.353

4.7 2.3 2.8 1.7 3.3 2.6 1.7 2.3 2.3 2.6 2.2 3.5 3.0 2.1 2.2 2.8 1.8 2.1

DS3 DS3 DS3 DS2 DS3 DS3 DS2 DS3 DS3 DS3 DS3 DS3 DS3 DS3 DS3 DS3 DS3 DS3 No Damage DS3

EQ-19

Kocaeli, Turkey

Arcelik

0.22

0.5

EQ-20

Chi-Chi, Taiwan

TCU045

0.474

3.49

4.6 SUMMARY Based on the results tabulated in the Table 4.4 obtained in this study, the following conclusions were gathered: 1. The analyzed bent is vulnerable for most of the simulated earthquakes, for like Northridge (recoding station: Beverly Hills), Northridge (recoding station: Canyon Country-WLC), Duzce (recoding station: Bolu), Superstition Hills (recoding station: El

Bridge Pier

Displacement Ductility,

Centro Imp. Co.) , Chi-Chi, Taiwan (recoding station: TCU045) would cause server

89

90

damage to the bridge bent. The displacement ductility demands for these earthquakes are more than 3.0.

CHAPTER 5 FRAGILITY OF RETROFITTED CONCRETE BRIDGE BENT

2. The ground motion of Landers (recoding station: Yermo Fire Station), Loma Prieta (recoding station: Capitola), Loma Prieta (Gilroy Array #3), Imperial Valley (Delta),

5.1 GENERAL

Manjil (recoding station: Abbar), Imperial Valley (recoding station: El Centro Array

An emerging tool in assessing the seismic vulnerability of highway bridges is the use of fragility

#11), Superstition Hills (recoding station: El Centro Imp. Co.), Superstition Hills

curves. Fragility curves describe the probability of a structure being damaged beyond a specific

(recoding station: Poe Road (temp)), Kobe (recoding station: Shin-Osaka), Cape

damage state for various levels of ground shaking. This, in turn, can be used for prioritizing

Mendocino (recoding station: Rio Dell Overpass), Kocaeli (recoding station: Duzce),

retrofit, pre-earthquake planning, and loss estimation tools. This is particularly useful in regions

Chi-Chi (recoding station: TCU045) would possess the bridge bent to degradation

of moderate seismicity, where bridge officials are beginning to develop retrofit programs, in

without collapse.

addition to conducting pre-earthquake planning. In light of the damage to bridges observed in

3. Due to Landers (recoding station: Yermo Fire Station) ground motion with PGA 0.24g

recent earthquakes, there is a significant need to perform adequate assessment of the

the bridge pier undergoes degradation (DS3) whereas a relatively higher PGA Landers

vulnerability of bridges and bridge networks prior to future seismic events. Fragility curves can

(recoding station: Coolwater) ground motion cause moderate damage (DS2). This

be either empirical or analytical. Empirical fragility curves are usually based on the reported

signifies that not only PGA but other factors such as velocity, energy of the ground

bridge damage from past earthquakes. Basoz et al. developed empirical fragility curves for the

motions are important factor for seismic response of the bridge bent.

bridge damage resulting from the 1994 Northridge, CA earthquake using logistical regression

4. It was found that ground motion Kocaeli (recoding station: Arcelik) with PGA 0.22g possess no harm to the bent.

analysis to account for uncertainties in the damage data. Shinozuka et al. used the maximum likelihood method to generate the empirical fragility curves from the observations of bridge damage in the 1995 Kobe earthquake. Billah et al. analyzed retrofitted multicolumn bridge bent for both near fault and far field earthquakes. Analytical fragility curves are developed through seismic response data from the analysis of bridges. The fragility analysis generally includes three major parts: (a) the simulation of ground motions, (b) the simulation of bridges to account for uncertainty in bridge properties, and (c) the generation of fragility curves from the seismic response data of the bridges. The seismic response data can be obtained from nonlinear time history analysis, elastic spectral analysis or nonlinear static analysis. The fragility curves are developed by performing nonlinear time history analyses for bridge pier. A set of 20 ground motion records, with varying magnitudes, distances, and peak ground accelerations, is used in the incremental dynamic analysis. The fragility curves are developed for various damage states for the pier.

5.2 GROUND MOTION FOR INCREMENTAL DYNAMIC ANALYSIS A suite of 20 near fault ground motions are used in this study to develop fragility curves for the as-built and retrofitted bridge bents. The far field ground motions were adopted for this analysis.

91

92

The characteristics of the earthquake ground motion records are presented in Table 5.1. All these

2.5

ground motions have very high PGA ranging from 0.24g to .0728g.

EQ1 EQ2 EQ3 EQ4 EQ5 EQ6 EQ7 EQ8 EQ9 EQ10 EQ11 EQ12 EQ13 EQ14 EQ15 EQ16 EQ17 EQ18 EQ19 EQ20 Average Average response spectrum

Table 5.1: Characteristics of the earthquake ground motion histories

Earthquake No

Name

Recording Station

PGAmax (g)

PGVmax (cm/s.)

EQ-1 EQ-2 EQ-3 EQ-4 EQ-5 EQ-6 EQ-7 EQ-8 EQ-9 EQ-10 EQ-11 EQ-12 EQ-13 EQ-14 EQ-15 EQ-16 EQ-17 EQ-18 EQ-19 EQ-20

Northridge Landers Northridge Landers Duzce, Turkey Loma Prieta Hector Mine Loma Prieta Imperial Valley Manjil, Iran Imperial Valley Superstition Hills Kobe, Japan Superstition Hills Kobe, Japan Cape Mendocino Kocaeli, Turkey Chi-Chi, Taiwan Kocaeli, Turkey Chi-Chi, Taiwan

Beverly Hills - Mulhol Yermo Fire Station Canyon Country-WLC Coolwater Bolu Capitola Hector Gilroy Array #3 Delta Abbar El Centro Array #11 El Centro Imp. Co. Nishi-Akashi Poe Road (temp) Shin-Osaka Rio Dell Overpass Duzce CHY101 Arcelik TCU045

0.416 0.24 0.4 0.283 0.7 0.53 0.3 0.56 0.2 0.51 0.4 0.36 0.5 0.45 0.2 0.385 0.3 0.353 0.2 0.474

58.95 51.5 43.0 26 56.4 35 28.6 36 26.0 43 34.4 46.4 37.3 35.8 38.0 43.8 59.0 70.65 17.7 36.7

Spectral Acceleration(g)

2

1.5

1

0.5

0 0.00

1.00

2.00 Period (sec)

3.00

4.00

Figure 5.1 Spectral acceleration of Earthquake ground motion records. Figure 5.2 shows different percentile of response spectral acceleration. A percentile is a measure used in statistics indicating the value below which a given percentage of observations in a group of observations fall. The 25th percentile is the value below which 25 percent of the observations may be found. The 25th percentile is also known as the first quartile (Q1), the 50th percentile as

Figure 5.1 shows the acceleration response spectra with 5% damping ratio of the recorded far

the median or second quartile (Q2), and the 75th percentile as the third quartile (Q3).

field ground motions. Figure 2b shows the different percentiles of acceleration response spectra with 5% damping ratio illustrating that the selected earthquake ground motion records are well describing the medium to strong intensity earthquake motion histories.

93

94

perimeter tie and to allow the formation of 135 degree hooks at the ends of the stirrups. The jacket was reinforced with 20–25mm vertical deformed bars with the same properties as the

1.6

vertical bars of the original column. The compressive strength of jacket concrete was 34 MPa,

Acceleration Response Spectra(g)

and the strain at peak stress was 0.002.

1.2

average

5.3 CHARACTERISTICS OF DAMAGE STATE

25th

In the seismic fragility analysis, different forms of EDPs are used to monitor the structural responses under earthquake ground motion and measure the damage state (DS) of the bridge

75th

components. Damage states for bridges should be defined in such a way that each damage state

0.8

indicates a particular level of bridge functionality. A capacity model is needed to measure the damage of bridge component based on prescriptive and descriptive damage states in terms of EDPs (FEMA 2003, Choi et al. 2004, Nielson 2005). Four damage states as defined by HAZUS (FEMA 2003) are commonly adopted in the seismic vulnerability assessment of engineering

0.4

structures, namely slight, moderate, extensive and collapse damages. Bridge piers are one of the most critical components, which are often forced to enter into nonlinear range of deformations under strong earthquakes. In this study, the displacement ductility of the bridge pier is adopted as

0 0.00

damage index (DI). Hwang et al, 2001 recommended four different damage states for bridge pier

0.50

1.00

1.50

2.00

2.50

3.00

3.50

4.00

Period(sec)

(Table 5.2) based on ductility limit. But retrofit affects the seismic response and demand of the bridge pier and the capacity as well. For the retrofitted bridge pier new limit states need to be defined. Limit states capacities for all the two retrofitted bridge bent are obtained by transforming

Figure 5.2: Percentiles of spectral acceleration of earthquake ground motion records

the ductility limit state proposed by Hwang et al, 2001 shown in table 5.2. The use of ductility

5.2 Details of Retrofitted Technique

limit for retrofitted RC columns is well documented in literature of Karthik et al, 2012, billah et al., 201

Among four different retrofit techniques, specifically, concrete jacketing, steel jacketing, FRP jacketing, and ECC jacketing, two retrofitted technique have been employed in this study. In this study, the CFRP composite jacket retrofitting technique has been implemented from Pantelides

TABLE 5.2: Damage/limit state of bridge components (Hwang et al, 2001, Kartik et al 2012, Billah et al, 2012) Damage State

Slight (DS=1)

Moderate (DS=2)

Extensive (DS=3)

Damage ((DS=3)

Cracking and Spalling

Moderate cracking and spalling

Degradation without collapse

Failure leading to collapse

and Gergely (2002), which has a tensile strength of 628 MPa, and ultimate axial strain of 10mm/m. The material is a carbon-fiber/ epoxy-resin composite with 48,000 fibers per tow

Bridge Component

Physical Phenomenon

As Built Bridge Pier CFRP Retrofitted Pier RCC Jacketing Retrofitted Pier

Displacement Ductility, Displacement Ductility, Displacement Ductility,

unidirectional carbon fibers. The thickness of the FRP jacketing was used to be 3.42 mm. Concrete jacketing is another potential technique for retrofitting the deficient multicolumn bridge bent. Applying full- or height concrete overlays to the face of an existing column can increase a column’s flexural strength. The new section must be well connected to the older one. Usually, a jacket thickness of 150 mm is used. This thickness is required to provide sufficient cover to the

95

96

(Shinozuka et al. 2000). The PSDM establishes a correlation between the engineering demand

5.4 INCREMENTAL DYNAMIC ANALYSIS In Incremental Dynamic Analysis (Hamburger et al., 2000; Vamvatsikos and Cornell, 2002), the structure is subjected to a series of non-linear time-history analysis of the increasing intensity (e.g. Peak ground motion acceleration is incrementally scaled from a low elastic response value

parameters (EDP) and the ground intensity measures (IM). In the current study, displacement ductility demand of retrofitted bridge bent was considered as the EDP, and the peak ground acceleration (PGA) was utilized as intensity measure (IM) of each ground motion record.

up to the attainment of a pre-defined post-yield target limit state. Incremental Dynamic Analysis

Two approaches are used to develop the PSDM: the scaling approach (Zhang and Huo 2009) and

(IDA) is a new methodology which can give a clear indication of the relationship between the

the cloud approach (Choi et al. 2004; Mackie and Stojadinovic 2004). In the scaling approach, all

seismic capacity and the demand. The analysis was carried out for the as-built and retrofitted

the ground motions are scaled to selective intensity levels and an incremental dynamic analysis

concrete bridge bent. The peak values of base shear are then plotted against their top

(IDA) is conducted at each level of intensity; however, in the cloud approach, un-scaled

displacement counterparts, for each of the dynamic runs, giving rise to the so-called dynamic

earthquake ground motions are used in the nonlinear time-history analysis and then a probabilistic

pushover or IDA envelop curves.

seismic demand model is developed based on the nonlinear time history analyses results. In the current study, the cloud method was utilized in evaluating the seismic fragility functions of the

5.5 FRAGILITY CURVE DEVELOPMENTS

retrofitted bridge bents. In the cloud approach, a regression analysis is carried out to obtain the

Fragility curve allows the evaluation of potential seismic risk assessment of any structure.

mean and standard deviation for each limit state by assuming the power law function (Cornell et

Fragility function describes the conditional probability i.e. the likelihood of a structure being

al. 2002), which gives a logarithmic correlation between median EDP and selected IM.

damaged beyond a specific damage level for a given ground motion intensity measure. The In this study, probabilistic seismic demand models are used to derive the fragility curves. The

fragility or conditional probability can be expressed as,

ground motions are scaled to selective intensity levels and an incremental dynamic analysis Fragility= P[LS|IM=y]

(5.1)

(IDA) is conducted at each level of the intensity. A regression analysis is carried out to obtain the mean and standard deviation for each limit state by assuming the power law function (Cornell et

Where, LS is the limit state or damage state of the structure or structural component, IM is the

al. 2002) , which gives a logarithmic correlation between median EDP and selected IM:

ground motion intensity measure and y is the realized condition of the ground motion intensity measure. In order to develop fragility curves different methods and approaches have been

EDP= a (IM)b or, ln (EDP) = ln (a) + b Ln (IM)

(5.2)

developed. Depending on the available data and resources, fragility functions can be generated empirically based on post-earthquake surveys and observed damage data from past earthquakes

Where, a and b are unknown coefficients which can be estimated from a regression analysis of

(Basoz et al. 1999; Yamazaki et al. 2000). However, limited damage data and subjectivity in

the response data collected from the nonlinear time history analysis. In order to create sufficient

defining damage states limit the application of empirical fragility curves (Padgett and DesRoches,

data for the cloud approach incremental dynamic analysis is carried out instead of nonlinear time

2008). In absence of adequate damage data, fragility functions can be developed using a variety

history analysis. The dispersion of the demand, βEDP| IM, conditional upon the IM can be

of analytical methods such as elastic spectral analyses (Hwang et al. 2000), nonlinear static

estimated from equation 5.3.

analyses (Shinozuka et al. 2000) and nonlinear time-history analyses (Hwang et al. 2001; Choi et al. 2004). In order to generate analytical fragility curves, structural demand and capacity needs to

√

∑

)

))

(5.3)

be modeled. In this study probabilistic seismic demand model (PSDM) was used to derive the analytical fragility curves using nonlinear time-history analyses of the retrofitted bridge bents. Although this is the most rigorous method, yet this is the most reliable analytical method

97

With the probability seismic demand models and limit states corresponding to various damage states, it is now possible to generate the fragilities using equation 5.3,

98

)

[

)

]

(5.4)

PSDM of as-built is of bridge pier is shown in Figure 5.4. The parameters of the PSDM are tabulated in Table 5.22 under near field earthquake ground motions.

)

)

)

(5.5)

y = 1.1931x + 1.5002 R² = 0.8011 (slight, moderate, extensive, collapse), a and b are the regression coefficients of the PSDMs and the dispersion component is presented in Equation 5.6. √

(5.6)

ln (Ductility)

ln(IMn) is defined as the median value of the intensity measure for the chosen damage state

Where, Sc is the median and βc is the dispersion value for the damage states of the bridge pier. The dispersion coefficient βc is used as describe by Karthik Ramanathan et al, 2012. The steps of fragility curve development shown in figure 5.3.

-5

-4

-3 ln (PGA)

-2

-1

0

Figure 5.4: PSDM of As-build concrete pier Table 5.3: PSDM parameter for two type of bridge pier Column Ductility Pier Condition

ln (a)

b

βEDP| IM

As-built

1.50

1.19

0.47

ln (Ductility)

y = 0.9539x + 0.9828 R² = 0.7629

-4

Figure 5.3: Fragility curve development 99

-3

-2 ln (PGA)

-1

Figure 5.5: PSDM of FRP Retrofitted concrete pier 100

0

Table 5.4: PSDM parameter for two type of Retrofitted bridge pier 1 Column Ductility

0.9

ln (a)

b

βEDP| IM

0.8

FRP Retrofitted

0.98

0.954

0.43

0.7

Propability of Damage

Pier Condition

y = 1.1893x + 1.3599 R² = 0.8054

0.6 0.5 0.4

FRP As-Built RCC Jacketing

0.3

0.2

ln (Ductility)

0.1 0 0

0.2

0.4

0.6

0.8

PGA(g) Figure 5.7: Fragility of as of concrete pier: Slight Damage -5

-4

-3 ln (PGA)

-2

-1

0 1 0.9

Figure 5.6: PSDM of Concrete Jacketed Retrofitted pier

FRP

Table 5.5: PSDM parameter for Concrete Jacketed Retrofitted pier Column Ductility Pier Condition

ln (a)

b

βEDP| IM

Concrete Jacketed Retrofitted

1.36

1.19

0.45

Propability of Damage

0.8 0.7

As-Built

0.6

RCC Jacketing

0.5 0.4

0.3 0.2 0.1

Plots of the fragility curves for two cases are shown in figure 5.7 to figure 5.10, which illustrated relative vulnerability of the retrofitted bridge bents over a range of far field Earthquake intensities

0 0

0.2

0.4 PGA(g)

0.6

and damage states. From figures it is evident that the analyzed concrete bent is capable to withstand moderate magnitude earthquake of PGA 0.2g without major damage.

101

Figure 5.8: Fragility of concrete pier: Moderate Damage 102

0.8

5.6 SUMMARY 1 This study investigated the performance of as-built bridge bent and the effectiveness of different

0.9

retrofitting techniques to the existing bent. To investigate the seismic vulnerability the as-build

Propability of Damage

0.8

FRP

and retrofitted bridge bents, a total of 20 earthquakes are utilized to evaluate seismic fragility of

0.7 0.6

As-Built

0.5

RCC Jacketing

the bent. The methodology for assessing the fragility of the retrofitted bridge bent includes the use of 2D nonlinear analytical models and time-history analysis and incorporation of the impact of different retrofits on fragility estimation. Through the process, the impact of a retrofit on the

0.4

PSDMs and the vulnerability of the bridge bent are evaluated. The fragility curves for bridge

0.3

bents are generated using the cloud approach for 20 far-field earthquake ground motion records.

0.2

Based on the analysis, the following conclusions can be drawn:

0.1 0 0

0.2

0.4 PGA(g)

0.6

0.8

1.

The numerical results in general show that the as-built bridge bents are more susceptible to seismic ground motions as compared with the retrofitted bent.

2. Among the two retrofitted technique used in the analysis concrete jacketing experiences more seismic vulnerability than FRP retrofitted technique under seismic ground motions.

Figure 5.9: Fragility of concrete pier: Extensive Damage

3. Analyses of the fragility curves reveal the effectiveness of a retrofit technique in mitigating probable damage due to far field earthquake. 4.

1 FRP

Propability of Damage

0.9

The fragility curves as obtained for the bridge bent can be used to estimate the potential losses incurred from earthquakes and it will help post-earthquake rehabilitation decision making, and hence selection of suitable retrofits techniques.

0.8

As-Built

0.7

RCC Jacketing

0.6 0.5 0.4

0.3 0.2 0.1 0 0

0.2

0.4 PGA(g)

0.6

Figure 5.10: Fragility of concrete pier: Collapse 103

0.8

104

CHAPTER 6 CONCLUSION

displacement pattern which helps to generate fragility curve. The fragility shows higher vulnerability to as-built concrete bent compared to retrofitted bents. By means of retrofitted measures the seismic vulnerability could be greatly reduced.

6.1 GENERAL

6. It can be inferred from the analysis that FRP retrofitted technique performed better

It was stated by Dr. Charles F. Richter that “Only fools, charlatans, and liars predict earthquakes” (Per Bak, How Nature Works). The uncertainty of the future earthquake makes it more difficult for the decision makes to assess the vulnerability for seismic activity. The major earthquake in future would cause seismic hazard to Flyover of Bahadarhat junction

under seismic event than concrete jacketing technique. The bridge bent needed to be retrofitted to escape from loss of life and serious injuries due to unacceptable bridge performance during large scale earthquake. Retrofitting with FRP would increase its capacity and the essential bridge function might maintain.

whereas the current code practice spectrum acceleration satisfies the capacity requirement. The bent shows lack of seismic capacity due to analyzed far field earthquake and leaves the flyover particularly vulnerable to earthquake. The result indicates that the bent is particularly vulnerable for earthquakes of more than PGA 0.2g. As awareness of the potential seismic

6.3 LIMITATION OF THE STUDY The main limitations of the current study are listed below:

threat needed to establish bridge retrofit programs and potential retrofit strategies for mitigate

Only one type of bridge bent geometry was considered, a total modeling of the bent with

seismic hazard. This study provides a through analytical investigation of seismic performance

bearing would be more effective for given earthquake vulnerability assessment ( Bhuiyan et.

of concrete bridge bent that illustrated the flyover would perform well under moderate

al., 2010)

earthquake within a PGA of 0.2g. The fragility curve of the analyzed bridge will help the decision maker to make retrofitting decision for the future earthquake event.

Only far 20 field earthquakes were considered for analysis. Future study may be conducted with near field earthquakes.

6.2 CONCLUSION

A further study using various bridges bent models with different sets of geometry/material

A thorough analysis of concrete bridge bent of the fly over is analyzed. Analytical method is

properties should be conducted for better understanding the contributions of other parameters

used to evaluate seismic performance of bridge bents to help for retrofitting and maintenance.

to the seismic fragility of a retrofitted bridge bent.

This study will lead to an effective solution for analyze concrete bridge bent system and its retrofitting work. 1. Based on the results obtained from static pushover and dynamic analysis the following conclusions are drawn: 2. It can infer from code shear strength and ductility evaluation that shorter piers are more susceptible to shear failure during an earthquake whereas the taller are more vulnerable to flexural failure. 3. These fragility curves can be used in determining the potential losses resulting from earthquakes, retrofitting prioritizations, and post-earthquake inspection decisions. 4. The dynamic analysis results shows variation in the results for different ground motions depends on not only on the PGA but also characteristics of ground motion records such as velocity, energy, etc. 5. The Incremental dynamic analysis gives the ability to generate different level of earthquake intensity of a ground motion to evaluate different shaking level and its

105

106

REFERENCES  A. H. M. Muntasir Billah, S.M.ASCE; M. Shahria Alam, M.ASCE; and M. A. Rahman Bhuiyan, “Fragility Analysis of Retrofitted Multicolumn Bridge Bent”, Subjected to NearFault and Far-Field Ground Motion”, Journal of Bridge Engineering, ASCE, October, 2013  A H M Muntasir Billah and M Shahria Alam, “Development of fragility curves for retrofitted multi-column bridge bent subjected to near fault multi-column bridge bent subjected to near fault ground motion”, 15th WCEE, 2012  Baker J and Cornell A C (2006), “Which Spectral Acceleration Are You Using?”, Earthquake Spectra, Vol. 22, No. 2.  Basoz, N. and Kiremidjian, A. S. (1999). “Development of empirical fragility curves for bridges.” Technical Council on Lifeline Earthquake Engineering Monograph Proceedings of the 1999 5th U.S. Conference on Lifeline Earthquake Engineering: Optimazing PostEarthquake Lifeline System Reliability, Aug 12-Aug 14 1999, (16), 693–702.  Basoz, N. and Mander, J. (1999). “Enhancement of the Highway Transportation Module in HAZUS.” Report No. National Institute of Building Sciences.

 Cornell, A. C., Jalayer, F., and Hamburger, R. O. (2002). “Probabilistic Basis for 2000 SAC Federal Emergency Management Agency Steel Moment Frame Guidelines.” Journal of Structural Engineering, 128(4), 526–532.  DesRoches, R., Leon, R. T., Choi, E., and Pfeifer, T. (2000). “Seismic Retrofit of Bridgesin Mid-America.” 16th US-Japan Bridge Engineering Workshop,  EERI, “Northridge Earthquake Reconnaissance Repost,: Earthquake Spectra, Special Supplement to Vol. 11, Feb 1995, 116 pp.  Ferracuti B and Savoia M (2005), “Cyclic Behaviour of Frp-wrapped Columns Under Axial and Flexural Loadings”, Proceedings of the International Conference on Fracture, Turin, Italy.  Filippou F C, Popov E P and Bertero (1983), “Effects of Bond Deterioration on Hysteretic Behaviour of Reinforced Concrete Joints,” Report EERC 83-19, Earthquake Engineering Research Center, University of California, Berkeley.  Hwang, H., Liu, J. B., and Chiu, Y. H. (2001). “Seismic fragility analysis of highway bridges.” Rep. Project MAEC RR-4, Mid-America Earthquake Center, Urbana, IL.

 Boore, D. M. (1983). “Stochastic Simulation of High-Frequency Ground Motions Basedon Seismological Models of the Radiated Spectra.” Bulletin of the Seismological Society of America, 73(6), 1865–1894.

 I.G. Buckle, 1996 “Overview of seismic design methods for bridges in different countries and future directions”, Eleventh World Conference on Earthquake Engineering

 Bostrom, A., Turaga, R. M. R., and Ponomariov, B. (2006). “Earthquake Mitigation Decisionsand Consequences.” Earthquake Spectra, 22(2), 313–327.

 Jamie E. Padgett,and Reginald DesRoches,” Methodology for the development of analytical fragility curves for retrofitted bridges”, Earthquake Engng Struct. Dyn. 2008; 37:1157–1174

 Bruneau, M. (1998). “Performance of steel bridges during the 1995 Hyogoken-Nanbu (Kobe, Japan) earthquake - a North American perspective.” Engineering Structures,20(12), 1063.

 Karthik Ramanathan, Reginald DesRoches and Jamie E Padgett (2012), “A Comparison of Pre-and Postseismic Design Considerations in Moderate Seismic Zones Through the Fragility Assessment of Multispan Bridge Classes”.

 Caltrans, Bridge Design Specification, California Department of Transportation, Sacramento, Calif, 1993  Chang, S. E., Shinozuka, M., and Moore, James E., I. (2000). “Probabilistic earthquake scenarios:Extending risk analysis methodologies to spatially distributed systems.” Earthquake Spectra, 16(3), 557.  Choi, E. (2002). “Seismic Analysis and Retrofit of Mid-America Bridges.” Report no., Georgia Institute of Technology.

 Madas P and Elnashai A S (1992), “A New Passive Confinement Model for Transient Analysis of Reinforced Concrete Structures,” Earthquake Engineering and Structural Dynamics, Vol. 21, pp. 409-431.  Mander J B (1999), “Fragility Curve Development for Assessing the Seismic Vulnerability of Highway Bridges”, Report No. 99-SP01, MCEER.

 Chai, Y.H.,M.J.N. Priestley and F.Seible, “Seismic Retrofit of Circular Bridge Columns for Enhanced Flexural Performance,” ACI Structural Journal, Vol. 88, September/October 1991, PP. 572-584.

 Menegotto M and Pinto P E (1973), “Method of Analysis for Cyclically Loaded R C Plane Frame Including in Geometryand Non-Elastic Behavior of Elements Under Combined Normal Force and Bending, Symposium on the Resistance and Ultimate Deformability of Structures Acted on by Well Defined Repeated Loads”, International Association for Bridge and Structural Engineering, pp. 15-22, Zurich.

 Choi, E., DesRoches, R., and Nielson, B. (2004). “Seismic Fragility of Typical Bridges in Moderate Seismic Zones.” Engineering Structures, 26(2), 187–199. Cimellaro, G. P. and Domaneschi, M. (2006). “Reliability of a Cable Stayed Bridge.” 8th U.S. National Conference on Earthquake Engineering, San Francisco, California.

 Nielson B (2005), “Analytical Fragility Curves for Highway Bridges in Moderate Seismic Zones”, Ph.D. Thesis, Georgia Institute of Technology.

 Nielson B (2005). “Personal Communication: Analytical Fragility Curves for Highway Bridges in Moderate Seismic Zones”.  Per Bak, How Nature Works: The Science of Self-Organised Criticality, New York: Copernicus, 1996.  Park Y J and Ang A H S (1985), “Mechanistic Seismic Damage Model for Reinforced Concrete”, ASCE Journal of Structural Engineering, Vol. 111, No. 4, pp. 722-39.  Pinho R, Casarotti C and Antoniou S (2007), “A Comparison of Single-run Push Over Analysis Technique for Seismic Assessment of Bridges”, Earthquake Engineering Structural. Dynamics, Vol. 36, pp. 1347-1362.  Priestley, M.J.N., F.Seible, and G. Macrae, The Kobe Earthquake of January 17, 1995: Initial Impressions from a Quick Reconnaissance, Structural Systems Research Project, Report SSRP-95/03, University of California, Sam Diego, February, 1995, 71pp.  Priestley, M.J.N., F.Seible, and C.M. Uang,”The Northbridge Earthquake of January 17, 1994: Damage Analysis of Selected Bridges, Structural Systems Research Project, Report SSRP-94/03, University of California, Sam Diego, February, 1994, 260pp.  Spoelstra M and Monti G (1999), “FRPConfined Concrete Model,” Journal of Composites for Construction, ASCE, Vol. 3, pp. 143-150.  Vamvatsikos D and Cornell C A (2002), “Incremental Dynamic Analysis”, Earthquake Engineering and Dynamics, Vol. 31, No. 3, pp. 491-514.  Yankelevsky D Z and Reinhardt H W (1989), “Uniaxial Behaviour of Concrete in Cyclic Tension,” Journal of Structural Engineering, ASCE, Vol. 115, No. 1, pp. 166-182.